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Full text of "Field study of pile group action"

no. 
FHW.A- 
RD- 
31-00 2 



?3 it Ztf5 



te rt No. FHWA/RD-81/002 

66? 

.A3 



// 



ID STUDY OF PILE GROUP ACTION 



March 1981 
Final Report 




^"o* 1. 



Document is available to the public through 
the National Technical Information Service, 
Springfield, Virginia 22161 



Prepared for 

FEDERAL HIGHWAY ADMINISTRATION 

Offices of Research & Development 

Materials Division 

Washington, D.C. 20590 



FOREWORD 



This report presents the results of a comprehensive research 
program to investigate the behavior of a vertically loaded 
pile group in stiff clay. The study included the development 
and verification of a mathematical model for pile group analysis, 
extensive pile and ground instrumentation, and 17 pile load tests 
to failure on single piles and pile groups. 

The work reported resulted from FCP Project 4H Study, "A Field 
Study of Pile Group Action," conducted by Raymond International 
Builders, Inc. The research was performed under DOT-FH- 11-9526 
during the period October 1, 1978, to November 18, 1980. 

In addition to the final report supporting documents are 
available upon request in the form of appendices to the final 
report, an interim report which describes the mathematical model 
chosen for detailed analysis and presents a priori analysis of 
group behavior, and a report on the analysis of dynamic 
measurements taken during driving the 11 test piles. 

Copies of the final report are being distributed by the Materials 
Division, Office of Research, to other researchers and to 
appropriate members of the FCP Project 4H team. 



Charles F. ScT^fey 
Director, Office of Research 
Federal Highway Administration 



NOTICE 

This document is disseminted under the sponsorship of the Department 
of Transportation in the interest of information exchange. The United 
States Government assumes no liability for its contents or use thereof. 

The contents of this report reflect the views of the authors who are 
responsible for the facts and the accuracy of data presented herein. 
The contents do not necessarily reflect the official view of the 
Department of Transportation. This report does not constitute a 
standard, specification, or regulation. 

The United States Government does not endorse products or 
manufacturers. Trade or manufacturers' names appear herein only 
because they are considered essential to the object of this document. 



f r ^1 



'/*/>' 



Technical Report Documentation Page 



1. Report No. 



FHWA/RD- 81/002 



2. Government Accession No. 



3. Recipient's Catalog No. 



%9 






4. Title and Subtitle 



FIELD STUDY OF PILE GROUP ACTION 



5. Report Date 

March 1981 



6. Performing Organization Code 



8. Performing Organization Report No. 



7. Author's) 

M. W. O'Neill, R. A. Hawkins, and L. J. Mahar 



9. Performing Organization Name and Address 

RAYMOND INTERNATIONAL BUILDERS, INC. 
2801 SOUTH POST OAK RD 
HOUSTON, TEXAS 77027 



10. Work Unit No. (TRAIS) 

FCP 34H2-012 



11. Contract or Grant No. 

DOT-FH- 11-9526 



12. Sponsoring Agency Name and Address 

OFFICES OF RESEARCH AND DEVELOPMENT 
FEDERAL HIGHWAY ADMINISTRATION 
U.S. DEPARTMENT OF TRANSPORTATION 
WASHINGTON , D.C. 20590 



13. Type of Report ond Period Covered 

Final Report 



14. Sponsoring Agency Code 

M/0671 



15. Supplementary Notes 

FHWA Contract Manager: Mr. Carl Ealy (HRS-21) 

Principal Investigator: Dr. Michael W. O'Neill, University of Houston 

Project Manager: Mr. Richard A. Hawkins, Raymond Technical Facilities, Inc. 



16. Abstroct 

This report is the final report for a study involving the static vertical load test- 
ing of a full scale, instrumented pile group. The test group consisted of nine pipe 
piles instrumented for settlement, load transfer, pore pressures, total pressures 
and inclination. Two similarly instrumented reference (control) piles were also in- 
stalled. Two smaller subgroups within the main group were also tested, and uplift 
tests were conducted on several of the individual piles. The soils at the test site 
consisted of clays that were overconsolidated by desiccation. It was determined that 
the efficiency of the main group and of the subgroups was essentially unity. Set- 
tlement ratios in the working load range were found to vary from about 1 . 2 to about 
1.7, depending on the number of piles that were loaded. Failure was observed to be 
by plunging of the individual piles . Unit side load transfer varied essentially 
linearly with depth. Some dependence of load transfer patterns on residual stresses 
that remained after driving the piles was observed. The measured behavior of the 
group and subgroups was modeled by the "hybrid" algorithm, by means of Program 
PILGP1 , which was developed for this study and documented in Appendixes A and B. 
Good agreement between computed and measured results were achieved when the unit 
load transfer curves from the reference piles were^sedand when the soil modulus 
of deformation was appropriately adjusted to account forpTTe"¥gfiff#9*pement of the 
soil and the presence of very small strains in the| maSr^j&ST&tEMT f^uild the group. 

TRANSPORTATION 
A description of the mathematical model, the rationale for its selection and a 

prior analysis of group behavior is presented in|FHV|a/]MH81/0pi^ Ab analysis of 

dynamic measurements taken during driving the 11 test piles is"'xn FfljwA/RD-81/009. 

Analyses and data obtained during the conduct of jthis study are in Appendixes A-F, 

FHWA/RD-81/003-008. J LIBRARY { 



17. Key Words 

Piles, Clay soils, Pile groups, 
computer model, pile driving 



18. Distribution Statement 

This document is available to the public 
through the National Technical Information 
Service, Springfield, VA. 22161 



19. Security Clossif. (of this report) 

Unclassified 



20. Security Classif. (of this page) 

Unclassified 



21. No. of Pages 

217 



22. Price 



Form DOT F 1700.7 (8-72) 



Reproduction of completed! page authorized 



CONVERSION FACTORS, U. S. CUSTOMARY TO METRIC (Si) 
UNITS OF MEASUREMENT 



U. S. customary units of measurement used in this report can "be con- 
verted to metric (Si) units as follows: 



Multiply 



Angstroms 

inches 

feet 

miles (U. S. statute) 

square inches 

square feet 

cubic feet 

cubic yards 

grams 

pounds (mass) 

tons (2000 pounds) 

pounds (mass) per cubic 
foot 

pounds (mass) per cubic 
yard 

pounds (force) 

pounds (force) per 
square inch 

pounds (force) per 
square foot 

miles per hour 

degrees (angle) 

Fahrenheit degrees 



Jy_ 



To Obtain 



0.0000001 (10" T ) 

2.5k 

0.30U8 

1.6093UU 

0.00061+516 

0.0929030U 

0.02831685 
O.T6I+55I+9 

0.001 

0.1*53592^ 

907.181+7 
16.0181+6 



0.59327631 

U.i;U8222 
689M57 

1+.882I+28 

1.6093M* 

0.0171+5329 

5/9 



millimetres 

centimetres 

metres 

kilometres 

square metres 

square metres 

cubic metres 

cubic metres 

kilograms 

kilograms 

kilograms 

kilograms per cubic 
metre 

kilograms per cubic 
metre 

newtons 

pascals 

kilograms per square 
metre 

kilometres per hour 

radians 

Celsius degrees or 
Kelvins* 



* To obtain Celsius (C) temperature readings from Fahrenheit (F) read- 
ings, use the following formula: C = (5/9)(F - 32). To obtain 
Kelvin (K) readings, use: K = (5/9)(F - 32) + 273.15. 



u 



TABLE OF CONTENTS 

Page 

Summary xi 

Chapter 1 - Test Pile Installation 1 

Introduction 1 

Soil Conditions at Test Site 2 

Test Piles 4 

Ground Instrumentation 11 

Loading and. Testing Sequence 11 

Pile Driving 13 

Soil Displacements During Driving 25 

Pore Water Pressures 27 

Assessment of Soil Disturbance 34 

As-Driven Locations of Piles 34 

Residual Loads Developed in Piles Due to Driving .... 40 

Chapter 2 - Pile and Soil Performance Under Load 45 

General 45 

Load- Settlement Behavior 45 

Distribution of Loads to Piles 67 

Variation of Capacity with Time 67 

Settlement Ratios 76 

Induced Settlements 78 

Efficiencies 81 

Pore Water, Total, and Effective Pressures 

Developed During Load Tests 89 

Ground Movements During Tests 112 

Chapter 3 - Load Transfer 127 

General 127 

Load Transfer Patterns for Reference Piles and for 

Group Piles by Position 127 

Apparent Peak Load Transfer by Soil Layer 141 

Progressive Failure Patterns 143 

Effects of Residual Stresses on Load Transfer 143 



in 



TABLE OF CONTENTS (Continued) 

Page 

Unit Load Transfer Curves 151 

Load Transfer Correlations 159 

Variability of Load Transfer 165 

Load Transfer Correlation Factors at Pile Tips 165 

Chapter 4 - Reanalysis of Performance 

Using Hybrid Model 169 

Introduction 169 

Tests Modeled 170 

Geometric Inputs 170 

Loadings 174 

Structural Properties 174 

Soil Inputs 174 

Results - Single Pile 180 

Results - Pile Groups 180 

Observations 186 

Chapter 5 - Recommendations for Future Study . ..... 194 

Appendix A 
Appendix B 
Appendix C 
Appendix D 
Appendix E 
Appendix F 



IV 



LIST OF FIGURES 



Figure Page 

A. Calibration of Piles xvii 

B . Driving of Test Piles . . . . . . . ■ . . . . . .xvii 

C. Pile Group Before Installing Cap . xviii 

D. Reaction and Reference Frames xviii 

E. Jacks and Load Cells xix 

F. Electronic Data Acquisition System xix 

1.1 Stratigraphy at Test Site 3 

1.2 Indicated Shear Strengths 5 

1.3 Indicated Young's Moduli 6 

1.4 At Rest Earth Pressure Coefficient and OCR 

Variation 7 

1.5 Instrumentation of Test Piles (Schematic) ...... 9 

1.6 Ground Instrumentation Plan 10 

1.7 Ground Instrumentation (East Elevation) 12 

1.8 Approximate Locations of All Test and 

Reaction Piles and Sequence of Driving 15 

1.9 Force-Time Traces for One Blow on Piles 2 and 4 . . 19 

1.10 Force-Time Traces for One Blow on Piles 1 and 5 . . 20 

1.11 Measured and Calculated Pile Forces and Wave 

Equation Model 22 

1.12 Measured and Computed Force-Time Relationships 

for Pile 4 23 

1.13 Observed Soil Movements During Pile 

Installation 26 

1.14 Observed Soil Pore Water Pressures Preceding 

Pile Installation 28 

1.15 Pore Water Pressure vs. Time at 19 ft (5.8 m) 

Depth 29 

1.16 Pore Water Pressure vs. Time at 34 ft (10.4 m) 

Depth 30 

1.17 Pore Water Pressure vs. Time at 50 ft (15.3 m) 

Depth 31 

1.18 Measured Effective Earth Pressure Coefficients 

Against Piles Four Days After Installation 35 

1.19 Pore Pressure Profile Section Lines 36 

1.20 Ground Pore Pressure Profiles: Before and at 
Conclusion of Driving 37 

1.21 As-Constructed Locations and Alignments of 

Group Piles 38 

1.22 As-Constructed Locations and Alignments of 

Reference Piles 41 

1.23 Residual Loads in Reference Piles After Driving ... 42 

1.24 Residual Loads in Typical Group Pile After 

Installation 43 



LIST OF FIGURES (Continued) 

Figure Page 

1.25 Comparison of Residual Loads (Per Pile) in 

Reference and Group Piles 44 

2.1 Reference Pile Load- Settlement Relationships 

for Test No. 1 46 

2.2 Reference Pile Load- Settlement Relationships 

for Test No. 2 48 

2.3 Reference Pile Load- Settlement Relationships for 

Test No. 3 49 

2.4 Cumulative Load- Settlement Curve for Reference 

Pile Tests 50 

2.5 Load- Settlement Relationships for 9-Pile Group, 

Test 1 51 

2.6 Cap Movements at Approximately One-Half of 

Failure Load, Test 1 52 

2.7 Cap Movements at Approximately 90 Percent of 

Failure Load, Test 1 53 

2.8 Cap Movements at Failure, Test 1 54 

2.9 Cap Movements Upon Removal of Load, Test 1 ... 55 

2.10 Load- Settlement Curves for Individual Piles 

on North Row, Test 1 .57 

2.11 Load- Settlement Curves for Individual Piles on 

Center Row, Test 1 .58 

2.12 Load- Settlement Curves for Individual Piles on 

South Row, Test 1 59 

2.13 Load- Settlement Curves for Second and Third 

9-Pile Group Tests 60 

2.14 Cumulative Load- Settlement Curve for 9-Pile 

Group Tests 61 

2.15 Load- Settlement Relationships for Subgroup Tests . . 63 

2.16 Normalized Load- Settlement Relationships 64 

2 . 17 Butt Load-Uplift Relationships . . 65 

2.18 Tip Load-Uplift Relationships 66 

2.19 Load Distribution to Pile Heads-Subfailure; 

Group Test 1 68 

2.20 Load Distribution to Pile Heads-Failure 

and Unloaded; Group Test 1 69 

2.21 Variation in Peak Capacity with Time 75 

2.22 Settlement Ratios for Nine-Pile Group Tests .... 77 

2.23 Measured and Theoretical Settlement Ratios for 

Nine-, Five-, and Four-Pile Group Tests 79 

2.24 Settlements in Unloaded Piles Versus Load Per 
Loaded Pile (above); Settlement Differences in 
Corner Piles Between 5- and 4-Pile Subgroup 

Tests (below) 83 

2.25 Pore and Total Pressure Changes on Piles at 9-Foot 

(2.7 m) Depth 91 

2.26 Pore and Total Pressure Changes on Piles at 

19-Foot (5.8 m) Depth 92 



VI 



LIST OF FIGURES (Continued) 
Figure Page 

2.27 Pore and Total Pressure Changes on Piles at 

34-Foot (10.4 m) Depth 93 

2.28 Pore and Total Pressure Changes on Piles at 

41-Foot (12.4 m) Depth 94 

2.29 Horizontal Variation in Pore Pressure in Soil on 

Pile 1 Prior to Reference Tests . 96 

2.30 Horizontal Variation in Pore Pressure in Soil 

and on Group Piles Prior to 9-Pile Tests 97 

2.31 Horizontal Variation in Pore Pressure in Soil and 

on Group Piles Prior to Subgroup Tests 98 

2.32 Horizontal Variation in Pore Pressure During 

Reference Pile Tests 99 

2.33 Horizontal Variation in Pore Pressure During 

9-Pile Tests 100 

2.34 Horizontal Variation in Pore Pressure During 

Subgroup Tests 101 

2.35 Vertical Variation in Pore Pressure on 

Reference Pile 1 103 

2.36 Vertical Variation in Average Pore Pressure on 

Group Piles; 9-Pile Tests 104 

2.37 Vertical Variation in Average Pore Pressure 

on Group Piles; Subgroup Tests 105 

2.38 Vertical Variation in Pore Pressure on Piles 

During Uplift Tests 107 

2.39 Vertical Variation in Total Lateral Pressure 

on Reference Pile 1 108 

2.40 Vertical Variation in Average Total Lateral 

Pressure on Group Piles (2,3,4,5) . . . ... .109 

2.41 Vertical Variation in Average Total Pressure 

on Piles in Subgroup Tests (2,3,5) 110 

2.42 Vertical Variation in Total Pressure for 

Uplift Tests Ill 

2.43 Vertical Variation in Lateral Effective Stress 

on Reference Pile 1 113 

2.44 Vertical Variation in Average Lateral Effective 

Stress on Group Piles (2,3,4,5) 114 

2.45 Vertical Variation in Average Lateral Effective 

Stress in Subgroup Tests (Piles 2,3,5) 115 

2.46 Vertical Variation in Lateral Effective Stress 

for Uplift Tests 116 

2.47 Surface Soil Movements Near Pile 1; Reference 

Tests 118 

2.48 Surface Soil Movements for 9-Pile Group Tests . . .119 

2.49 Surface Soil Movements for Average of Subgroup 

Tests 120 

2.50 Soil Movements; 300 inch (7.6 m) Depth: 9-Pile 

Group Tests 121 



Vll 



LIST OF FIGURES (Continued) 



Figure 

2.51 

2.52 

2.53 

2.54 

3.1 

3.2 

3.3 

3.4 

3.5 

3.6 

3.7 

3.8 

3.9 

3.10 

3.11 
3.12 
3.13 
3.14 
3.15 
3.16 
3.17 

3.18 
3.19 
3.20 
3.21 
3.22 



Page 



Soil Movements; 300 inch (7.6 m) Depth; Average 

of Subgroup Tests 122 

Soil Movements; 516 inch (13.1 m) Depth; 9-Pile 

Group Tests 123 

Soil Movements; 516 inch (13.1 m) Depth; Average 

of Subgroup Tests 124 

Soil Movements; 600 inch (15.3 m) Depth; 

9-Pile Group Tests 125 

Load Distribution and f-d Diagrams 
Subfailure; 9-Pile Test 1 . . . 
Load Distribution and f-d Diagrams 
Failure; 9-Pile Test 1 . . . . 
Load Distribution and f-d Diagrams 
Subfailure; 9-Pile Test 2 . . . 
Load Distribution and f-d Diagrams 
Failure; 9-Pile Test 2 . . . . 
Load Distribution and f-d Diagrams 
Subfailure; 9-Pile Test 3 . . . 
Load Distribution and f-d Diagrams 

9-Pile Test 3 

Load Distribution and f-d Diagrams 

5-Pile Test 

Load Distribution and f-d Diagrams 

5-Pile Test 

Load Distribution and f-d Diagrams 

4-Pile Test 

Load Distribution and f-d Diagrams 



Failure ; 



Subfailure ; 



Failure ; 



Subfailure; 



Failure ; 



128 



129 



130 
131 



132 



133 



134 



135 



136 



4-Pile Test 137 

Progressive Failure in Reference Piles; Test 1 144 

Progressive Failure in Group Piles; 9-Pile Test 1 . . 145 

Residual Loads in Reference Piles; Test 1 146 

F-d Relationships for Reference Piles; Test 1. 147 

Average Residual Loads in Group Piles; Test 1 . 149 

Average f-d Relationships for Group Piles; Test 1 . . 150 
Average Apparent and Adjusted Load 
Distribution Diagrams at Failure for Reference 

Piles 152 

Average Apparent and Adjusted Load Distribution 
Diagrams at Failure for Group Piles 2,4,5 and 9 . 153 

F-z Curves; Soil Zones A and B; Reference 

Piles; Test 1 155 

F-z Curves; Soil Zones C and D; 

Reference Piles; Test 1 156 

F-z Curves; Soil Zones A and B; 

Group Piles; Test 1 157 

F-z Curves; Soil Zones C and D; 

Group Piles; Test 1 158 



Vlll 



LIST OF FIGURES (Continued) 

Figure Page 

3.23 Q-z Curves for Test 1; Reference Piles 

(above); Group Piles (below) 160 

4.1 Pile Head Coordinates for PILGP1 Analysis 171 

4.2 Direction Angles 172 

4.3 Jack Coordinates 175 

4.4 F-z Curves for PILGP1 Input 176 

4.5 Q-z Curve for PILGP1 Input 177 

4.6 Computed and Measured Mean Pile Head 

Load Settlement Curves; Reference Piles; Test 1; 

PILGP1 181 

4.7 Computed and Measured Mean Distribution 

of Load; Reference Piles; Test 1; PILGP1 182 

4.8 Measured and Computed Load- Settlement Curves; 

9-Pile Test 1 183 

4.9 Measured and Computed Settlement Ratios; 

9-Pile Test 1 185 

4.10 Measured and Computed Distributions of Loads 

Along Piles; 9-Pile Test 1; Load = 581.4 K . . . .188 

4.11 Measured and Computed Distribution of Loads Along 

Piles; 9-Pile Test 1; Load = 1274.7 K 189 

4.12 Measured and Computed Load- Settlement 

Curves; Subgroup Tests 190 

4.13 Measured and Computed Distributions of Loads Along 
Piles; 5-Pile Test; Load = 278.9 K 191 

4.14 Measured and Computed Distribution of Loads 

Along Piles; 4-Pile Test; Load = 287.6 K 192 



IX 



LIST OF TABLES 
Tables Page 

A. Summary of Gross Test Results xiv 

B. Summary of Load Distribution Data: 

First Nine-Pile Group Test Series xv 

1.1 Chronology of Major Field Events 14 

1.2 Pile Driving Data: Blow Counts in Blows/Foot ... 17 

1.3 Computed Peak Pile Forces During Driving 

As Function of Side Damping , J 24 

1.4 Pile Head and Jack Coordinates tor 9-Pile 

Test No. 1 39 

2.1 Distribution of Loads to Pile Heads: 9-Pile 

Group Test 70-72 

2.2 Distribution of Loads to Pile Heads: Subgroup 

Tests 73-74 

2.3 Settlement Ratios for Pile Tips for Test 1 80 

2.4 Induced Settlements 82 

2.5 Summary of Failure Loads and Efficiencies for 

9-Pile Group Tests 84 

2.6 Summary of Failure Loads and Efficiencies for 

Subgroup Tests 86 

2.7 Overall Efficiency by Geometric Position and 

Average Shaft and Tip Efficiencies 87 

2 . 8 Overall Pile Group Efficiencies for Test Program ... 88 

3.1 Variation of Depth of Median Side Load Transfer 

Among Tests 140 

3.2 Variation of Peak Unit Side Load Transfer Based 

on Pretest Zeros 142 

3.3 Adjusted and Unadjusted Peak Load Transfer in psf 

by Soil Layer for Compression and Uplift Tests . . . 154 

3.4 Side Resistance Correlation Factors, 9-Pile 

Test 1 161 

3.5 Interpreted Peak Cohension (c) and Angle of 
Internal Friction (<{>) Values Used for Load Transfer 
Correlations 163 

3.6 Factors (6) for Computing a Cor elation for Individual 
Piles and Layers, 9-Pile Test 1 ,.-.. . 166 

3.7 Average End Bearing Capacity Factors for 

Reference and Group Piles 167 

4.1 Pile Geometry for PILGP1 Reanalysis 173 

4.2 F-z Curves for PILGP1 Solutions: Reanalyzed and 
Criteria 178 

4.3 Q-z Curves for PILGP1 Solutions 179 

4.4 Distribution of Loads to Piles From PILGP1 

Analyses 187 



Summary 

The purposes of the study described in this report were to obtain 
field data that are useful in interpreting fundamental phenomena that 
control the behavior of groups of driven piles in overconsolidated clay 
and that are appropriate for verification of mathematical models for 
predicting the response of pile groups to applied vertical, static loads. 
In order to develop the data, eleven instrumented full-sized pipe piles 
were driven: two as isolated, reference piles and nine in a square 
group. These piles were load tested to failure at various times after 
driving as individual piles, as a nine-pile group with a three -diameter 
spacing, and as five- and four-pile subgroups with variable spacings. 
A comprehensive in- situ and laboratory soil investigation program was 
undertaken, and piezometers and movement monuments were placed in 
the soil. 

A specific mathematical model, the "hybrid" model, was selected for 
detailed study. That model and other models are described in the 
Interim Report . A digital computer version of the hybrid model, 
Program PILGP1, was developed during the study and is described and 
documented in Appendix A and Appendix B of this report. 

Gross field results are summarized succinctly in Tables A and 
B. The term "SP" in the former table refers to "single pile." 

The following principal observations were made: 

1. Significant pore water pressures were developed during 
driving adjacent to the reference (isolated) pile that was studied for 
this effect and in the soil within and surrounding the pile group. Pore 
pressures dissipated rapidly thereafter such that pore pressures were 
nearly hydrostatic within about 20 days after driving within the group 
and around the reference pile. The rapid pore pressure dissipation is 
thought to be due largely to the presence of a secondary structure 
network and to continuous sand partings in the soil layers and to the 
high coefficient of consolidation associated with the overconsolidated 
soils. Very small positive changes in pore pressure occurred during 
loading . 

2. The distribution of soil resistance along the piles during the 
driving process was dissimilar to that observed during static loading. 
During driving, essentially no dissipation of compression wave amplitude 
occurred over the top half of the piles at full penetration. This 
suggests that no shear stress transfer took place in the top half of the 
piles during driving, possibly resulting from a small annular space 
between piles and soil that may have developed because of lateral motion 
of the piles. Significant transfer of load occurred in this zone during 



XI 






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Xlll 



static testing, implying that the annular space may have been closed 
due to lateral soil expansion after driving. 

3. Effective lateral stresses measured in the piles several days 
after driving were greater than the in-situ lateral pressures, especially 
in the bottom halves of the piles. 

4. Shear strength of soil within the area of the pile group but 
some distance away from the immediate vicinity of the pile faces, as 
measured by the static cone resistance after testing, changed negligably 
from pre-drive strength. 

5. Observed butt heave on piles already in place in the group 
never exceeded 0.05 in. (1.3 mm) due to driving of the remaining 
piles. Maximum soil surface heave was approximately 1 in. (25.4 mm) 
adjacent to the 9-pile group, diminishing to 0.1 in. (2.54 mm) 28 ft. 
(8.54 m) from the center of the group. 

6. Small, variable residual loads were observed in the reference 
piles after installation. Residual loads were somewhat lower in the 
group piles at the same time. Static load testing to failure in both the 
group and reference piles produced a significant increase in residual 
loads which had to be considered in order to assess set-up effects. 

7. Appreciable "apparent" set-up occurred between the first and 
second sets of loads tests (lapse period of 62 days) in both the 
reference piles and group piles. This effect could not be attributed to 
increased side resistance caused by pore pressure reductions, which 
were very minor during the lapse period. Instead, they were deduced 
to be largely the result of increased tip capacity resulting from the 
effects of the earlier loading. 

8. The efficiency of the pile groups (9, 5, and 4 piles) was 
essentially 1.0 when the weight of the pile cap was included as applied 
load. Shaft efficiency was slightly less than 1.0, and tip efficiency was 
greater than 1.0. Since failure was essentially "brittle," shaft failure 
preceded tip failure, whereafter shaft relaxation occurred in both 
reference and group piles. Therefore, the peak capacity of any 
individual pile was slightly less than the sum of its peak side and tip 
capacities . 

9. Failure of the 9-pile group and of the subgroups was by 
plunging of the individual piles and not by failure of the groups as 
blocks. 

10. The settlement ratios at subfailure loads in the 9-pile group 
and in the subgroups were considerably lower than the settlement ratios 
predicted by elastic solid models when the soil modulus was constant or 
uniformly increasing with depth in a strained layer that is infinitely 
thick. Several factors could account for this effect, including the 



XIV 



reinforcement of the soil provided by the piles and the relatively strong 
influence of the stiffer soils that were present beneath the pile tips 
that were not modeled properly with the procedures considered. 

11. The mean measured shear strain amplitudes in the soil 
immediately adjacent to the group were very small at loads up to and 
including the failure load. This fact suggests that elastic modulus to 
be used in analytical models for predicting short-term settlement 
response of pile groups should be taken to correspond to very low 
strain amplitudes and should be measured in-situ whenever possible. 

12. The loads were relatively evenly distributed to the pile heads. 
Within the working load range in the 9 -pile group the center pile 
carried the least load and the corner piles carried the greatest load. 
These loads differed by about 10 per cent. Larger differences are 
generally predicted by elastic solid models. (See Interim Report . ) 

13. Progressive failure was experienced in the soil at the test 
site. In an individual group or reference pile, shaft failure progressed 
generally from the top and bottom of the pile toward the middle, 
beginning at an applied load of about 85 percent of plunging failure 
load. During the first group load test slightly eccentric loading caused 
tipping of the pile cap, which induced plunging failure in the northeast 
corner pile that then progressed to other piles in the group. In the 
other group and subgroup tests, where loading was more concentric, 
failure was essentially non-progressive among the piles. 

14. Graphs of peak developed unit side resistance versus depth 
showed a distinct trend toward linear increase in unit side resistance 
with depth. This fact, coupled with the small measured pore pressure 
changes observed during loading, suggests that a frictional or effective 
stress approach to assessment of shaft capacity is feasible for soils of 
the type in which the tests were conducted. 

15. The best overall direct correlations of load transfer, both at 
the tips and along the shafts of the piles, were with the in-situ static 
cone, although some variability in the side resistance correlation factors 
existed among the various soil layers. The general effective stress 
method (GESM) also yielded correlation factors near, but consistently 
below 1.0 (indicating unconservative predictions) below a depth of 10 
ft. (3.05 m). Correlation of measured maximum unit side load transfer 
with soil shear strength calculated from peak effective stress parameters 
measured in laboratory triaxial compression and from the measured 
lateral effective stresses on the piles were less accurate, due partially 
to the difficulties in interpreting total normal stresses (Item 17) and the 
questionable applicability of peak triaxial parameters to represent soil 
shear strength. Indirect correlation methods, such as the a method, 
yielded factors consistent with those obtained in numerous single pile 
tests in stiff, overconsolidated clay. 



XV 



16. No conclusions can be drawn with respect to the comparison 
of behavior of a reference pile tested according to the "quick test" 
method and one tested according to the "standard" one-hour load 
increment method used in this study. Apparently, the pile that was 
subjected to the quick test was enveloped by soil with higher 
undissipated pore pressures at the time the comparative tests were 
carried out than the pile subjected to the standard test, resulting in its 
unexpectedly lower capacity. This is speculated to be due to the result 
of driving the pile subjected to the quick test in a pilot hole that was 
partially filled with water. 

17. The instrumentation and data acquisition systems (described 
in detail in the Interim Report ) performed adequately, except that some 
ground piezometers dTd not function properly and the temperature 
sensitivity and small geometric irregularities in the total lateral pressure 
cells made interpretation of total pressure data difficult. In this 
regard, the total and effective stresses shown in Chapter 2 should be 
considered as representative of trends and not of exact values. 

18. Program PILGP1 satisfactorily replicated the behavior of the 
piles during the first 9-pile group test and the 5- and 4-pile subgroup 
tests. Unit load transfer curves from the first set of reference pile 
tests were used as inputs, and the soil was treated as incompressible 
(Poisson's ratio = 0.5) and given a Young's modulus about twice that 
measured by the self-boring pressuremeter in the soil at a level 
immediately below the pile tips. This value of Young's modulus also 
corresponds to an E/c of 1400, where c is the average undrained 
cohesion, as indicated by UU triaxial compression tests, between the 
ground surface and the depth of the pile tips. 

19. The experimental results obtained in this study are directly 
applicable only to small groups of moderately spaced driven displacement 
piles in soil of the type encountered at the test site. Application of 
the results to sites with other soil types or to larger groups or groups 
with more closely spaced piles in any type of soil must be done through 
sound judgment. One use of the hybrid model (or other mathematical 
model) would be to assist a designer in making that judgment. 

20. Stress contours in the soil in around the piles were not 
measured. However, displacements were measured at several points in 
the soil, and stresses induced in unloaded piles were also measured. It 
may be possible to use indirect analytical methods to infer approximate 
stress contours from these data, but such analyses are beyond the 
scope of this report. 

Photographes of several aspects of the field work are shown in 
Figs. A-F. 



XVI 




FIGURE A. CALIBRATION OF PILES 





4 


1 n 1 








■ ■ '' : Mi 


■ ffi* l ' v 








- B \ 






, 






. 






HE Tr 




1 


Km 













FIGURE B. DRIVING OF TEST PILES 



xvu 




FIGURE C. PILE GROUP BEFORE INSTALLING CAP 




FIGURE D. REACTION AND REFERENCE FRAMES 



XVlll 







FIGURE E. JACKS AND LOAD CELLS 




FIGURE F. ELECTRONIC DATA ACQUISITION SYSTEM 



xix 



Chapter 1. Test Pile Installation 

Introduction 

This report describes the results of measurements made during the 
installation and vertical load testing of eleven instrumented test piles in 
overconsolidated clay at a test site on the University of Houston Central 
Campus. The overall objectives of the study were: (1) to evaluate 
mathematical models for pile groups; (2) to choose one model to predict 
the behavior of a three by three group of piles that would be tested in 
over- consolidated clay; (3) to design a test program, based on the 
aforementioned mathematical solution, that would be capable of verifying 
the model; (4) to install and conduct load tests on the pile group and 
two reference (control) piles, acquiring such data as needed to check 
the mathematical model and other data relevant to the fundamental 
understanding of pile group response under vertical loading; (5) to 
remodel the load test using the chosen model in order to calibrate the 
model to the soil conditions encountered at the test site. 

The Interim Report for the study, dated March, 1979, was con- 
cerned with Items 1-3, above. Procedures for modeling the pile group 
mathematically have been described in the Interim Report . That report 
also contains details of the pile and soil instrumentation systems 
employed, the pile driving system, geotechnical conditions, testing 
procedures, and structural details of the reaction and reference systems 
and the cap. This report is concerned with Items 4 and 5. 

This report, while factual, necessarily reflects some degree of 
interpretation of data and phenomena by the authors. So that other 
investigators may conduct independent analyses, a complete set of raw 
and partially reduced data has been transmitted to the Federal Highway 
Administration, Offices of Research and Development, Washington, D.C. 
20590. Supporting documentation and data (principally in reduced 
form) for this report are contained in several appendices, as follows: 

(1) Appendix A: User's Guide for Program PILGP1. This 
digital computer program is the algorithm for the hybrid mathe- 
matical model chosen for use on this project. The hybrid model 
characterizes soil response by combining unit load transfer curves 
and elastic theory. The hybrid model is described in the User's 
Guide. 

(2) Appendix B: Documentation for Program PILGP1. This 
appendix contains flow charts, descriptions of primary variable 
names , etc . , used in Program PILGP1 . A complete listing of the 
FORTRAN IV computer code is also provided. 

(3) Appendix C: Geotechnical Investigation. Appendix C 
contains the geotechnical data for the site. (A distillation and an 
interpretation of the geotechnical conditions are given in Chapter 5 



of the Interim Report .) Information contained in this appendix 
includes results from static cone soundings obtained before and 
after driving the piles, pressuremeter tests, seismic crosshole 
tests, standard penetration tests, water level measurements, 
classification tests, torvane and hand penetrometer tests, moisture 
content and dry density tests, unconfined compression tests, 
undrained triaxial compression tests with pore pressure measure- 
ments, and drained direct shear tests. Locations of the test 
borings are noted, and detailed logs of boring are provided. 

(4) Appendix D: Detailed Graphical Presentation of Reduced 
Data. This appendix contains selected representative computer- 
produced graphs of load-settlement, load distribution, and load 
transfer data for all tests except the first test, which is described 
in detail in the main text. 

(5) Appendix E: Evaluation of Instrumentation. Appendix E 
describes in detail problems encountered with the various pile and 
soil instrumentation systems, documents transducer performance, 
describes pile calibration, and describes the procedure used for 
fitting strain gage data. Sources and evaluation of errors in 
deformation measurements are also considered. 

(6) Appendix F: Supplementary Information. Certain 
repetative supplementary reduced data, such as graphical 
presentation of cap motion, load distribution diagrams, and total 
and pore pressure tabulations are contained in Appendix F. 

Soil Conditions at Test Site 

While numerical descriptions of soil properties are given in both 
the Interim Report and in Appendix C of this report, a short 
description of the soil conditions at the test site is given here. 

Two principal geological formations were identified at the site: (1) 
the Beaumont formation, from the ground surface to a depth of 26 ft. 
(7.9 m); and (2) the Montgomery formation, sometimes described locally 
as the Upper Lissie formation, below a depth of 26 ft. (7.9 m). Both 
of these formations are deltaic Pleistocene terraces, with the underlying 
Montgomery formation having been deposited during the Sangamon 
Interglacial Stage and the Beaumont formation having been deposited 
during the Peorian Interglacial Stage. Both deposits consist primarily 
of clay that was preconsolidated by means of desiccation when the level 
of the nearby Gulf of Mexico was several hundred feet below its present 
level. 

The site topography is essentially flat. The stratigraphy consists 
of 1.5 ft. (0.5 m) of clay fill, below which is found approximately 6 ft. 
(2 m) of weathered Beaumont clay. A detailed stratigraphic schematic 
is shown in Fig. 1.1. Stratum C, the upper 4 ft. (1.2 m) of the 



20- 



30- 



o. 

UJ 

o 




SPT COUNT 
(BLOWS/ft) 

i — i — i — i — i 
10 20 3040 



40- 



50- 



60 L 



TRUE STRATIGRAPHY 



VERY STIFF GRAY AND TAN CLAY (CL-CH) 

STIFF GRAY AND TAN SANDY CLAY 

WITH SAND SEAMS (CD 
© STIFF TO VERY STIFF RED AND 
_ LIGHT GRAY CLAY(CH) 
® STIFF TO VERY STIFF LIGHT GRAY 

AND TAN SANDY CLAY WITH 

SAND POCKETS (CD 



<2> DENSE RED AND 
LIGHT GRAY SILT 
WITH CLAYEY SILT 



IDEAL STRATIGRAPHY 
(FOR CORRELATIONS ) 



ZONE ®- STRATUM (J) 
AND SAND LAYERS(ML) ZONE ®- STRATA <g> a ( 
VERY STIFF RED ZONE ©- STRATUM <g> 

AND LIGHT GRAY CLAY (UPPER) 

(CD ZONE©- STRATUM © 

(LOWER) 
ZONE ©- STRATA 

8 BELOV 



FIGURE 1.1. STRATIGRAPHY AT TEST SITE (1 ft = 0.305m) 



Montogomery formation, was noticeably softer than underlying soil. 
Note that the free water level in observation wells remained at 
approximately 7 ft. (2.1 m) below site grade throughout the study. 
Piezometric studies, described later in this chapter, indicate that this 
water level closely represents the piezometric head in all soils down to a 
depth of 50 ft. (15.3 m), which is 7 ft. (2.1 m) below the tips of the 
test piles. 

Figure 1.2 expresses the variation of shear strength as indicated 
by several methods. The static cone penetrometer values are averages 
based on three soundings made prior to pile installation. 
Inconsistencies among methods are readily apparent. Most are believed 
to be assocaited with the strong secondary structure of the Beaumont 
formation and of the relatively high sand content of the Montgomery 
formation (often present in small lenses and partings), which produce 
scatter in parameters obtained through laboratory tests and probably 
result in a bias toward unrepresentatively low laboratory strengths, 
especially in the sandy clay soil below a depth of 35 ft. (10.7 m), 
where the content of the sand increases to almost 50 per cent. Further 
explanation of the test results reported in Fig. 1.2 is given in the 
Interim Report . 

Since it will be necessary to refer to the Young's modulus of the 
soil during the mathematical analysis of the tests described in Chapter 
4, Fig. 1.3 is reproduced from the Interim Report to express the 
variation of Young's modulus with depth by several test methods. The 
moduli reported for the laboratory triaxial test are arbitrarily defined 
as secant moduli to the principal stress difference-major principal strain 
curve at twenty per cent of the peak principal stress difference. It is 
evident that the associated strain level in the triaxial text is much 
higher than that produced by crosshole seismic testing. Again, the 
variability of data is obvious. Part of this variability is associated with 
the definition of strain level at which the modulus is defined; part is 
also undoubtedly due to the effects of sampling disturbance, rendering 
the laboratory values unrepresentatively low. 

Figure 1.4, which was also extracted from the Interim Report , 
presents graphs showing the measured variations of in-situ at-rest 
earth pressure coefficient and overconsolidation ratio (OCR) with depth. 
Note that the at-rest earth pressure coefficient is very high in the 
depth range of 15 - 20 ft. (4.6 - 6.1 m). This range may have 
concided with an ancient evaporation surface. 

Test Piles 

The test piles were steel pipes, 10.75 in. (273 mm) in outside 
diameter, with wall thicknesses of 0.365 in. (9.27 mm). They were 
closed on the botton end with 1 in. (25.4 mm) thick steel boot plates 
and sealed at the top (after driving) with airtight cover plates through 



UNDRAINED SHEAR STRENGTH (ksf) 
2 4 6 8 10 




AVERAGE CONE TIP (D) 



PRESSUREMETER (▲) 



FIGURE 1.2. INDICATED SHEAR STRENGTHS (1 ft = 0.305 m; 
1 ksf =47.9 kN/m 2 ) 



YOUNG'S MODULUS (psi) 
5,000 10,000 15,000 20,000 40,000 60,000 80,000 100,000 

-WS 1 1 i 1 



10 



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I- 

Q_ 
Q 



30 



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50 




A 
\ 



\ 
\ 



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a\ 

\ 



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\ 
\ 
\ 



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NORMALIZED 

PARAMETERS (□) 



UU 
^TRIAXIAL 
COMPRESSION (•) 



PRESSUREMETER(A) 



-CROSSHOLE(A) 
(COMPUTED 

FROM SHEAR 

MODULUS ) 



60 L 



FIGURE 1.3. 



INDICATED 
1 psi 



YOUNG'S MODULI 
= 6.89 kN/rn 2 ) 



(1 ft = 0.305 m; 





r 



OVERCONSOLIDATION RATIO (OCR) 

2 4 

1 1 

2 



T 



Ko 



3 



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i 

Ql 
U 
Q 



30 



40 - 



c7 



OCR 



50 




A PRESSUREMETER 

• CONSOLIDATION (l-d) 

▼ Ko -CONSOLIDATION TRIAXIAL 

OCR 

O CONSOLIDATION (l-d) 

V HIGH-PRESSURE-CONSOLIDATION 
TRIAXIAL 



60 L 



FIGURE 1.4. AT REST EARTH PRESSURE COEFFICIENT AND OCR 
VARIATION (1 ft = 0.305 m) 



which dry nitrogen was passed in the periods between tests to preserve 
the integrity of the electrical instruments. 

Nine of the piles were driven in a square group, nominally three 
diameters on centers. These piles penetrated to a depth of 43 ft. (13.1 
m) and protruded 8.25 ft. (2.5 m) above the ground surface. The 
piles were rigidly connected to each other with a 4.25-ft.- (1.30 m) 
thick reinforced concrete cap. A 3.0-foot (0.92 m) free space was left 
between the bottom of the cap and the soil surface. Two other piles 
were driven as isolated reference (control) piles 12 ft. (3.66 m) to the 
south and north of the center of the 9-pile group, respectively. These 
piles were identical to the group piles except that they were not capped 
and extended only 3.3 ft. (1.0 m) above grade. 

The instrumentation carried by the various test piles is summarized 
schematically on Fig. 1.5. All eleven test piles contained full-bridge, 
precalibrated strain gage circuits placed at approximately 5 ft. (1.5 m) 
depth intervals. Before driving, each pile was subjected to an 
"exercising" procedure followed by calibration of each gage circuit to an 
axial load of 150 tons (1335 kN). Details of the calibration procedure 
are given in Appendix E. The second highest levels of strain gages in 
the group piles served as load transducers for measuring load 
distributed to the head of each pile, with the uppermost level (just 
beneath the base of the pile cap) serving as a backup load transducer 
in case of a malfunction of the primary transducer. The remaining 
strain circuits were used to measure load transfer between the piles and 
the soil. 

The pile numbering scheme is shown in Fig. 1.6. The group piles 
were designated with numbers 2-10, while the reference piles were 
assigned numbers 1 and 11. 

Load applied to the pile group and to the reference piles was also 
measured by a series of independent electronic load cells, and loads 
were also monitored by reading hydraulic jack pressures. All of the 
electronic instrumentation was monitored with a microcomputer-based 
data acquisition system, described in the Interim Report . 

Other pile instrumentation consisted of mechanical extensometers to 
serve as a backup to the strain gage system in the event of electronic 
system malfunction, lateral earth pressure (total and pore water 
pressure) cells at four locations on five of the test piles, and an 
inclinometer tube to be used after installation to determine the exact 
orientation of the piles in the ground. A separate strain gage system, 
consisting only of single gages, was installed and monitored during the 
driving of four of the piles (Nos. 1,2,4, and 5). This "dynamic" system 
was used to obtain force-time traces for the purpose of comparing the 
results of an E.A.L. Smith wave equation model analysis with measured 
behavior. This wave equation analysis, described later in this chapter, 
was used to evaluate the appropriate values of side damping and distri- 
bution of static resistance for isolated and group piles at this test site. 

8 



EXTENSOMETER SENSOR 
HEADS INSTALLED 
AFTER DRIVING 



3/8" STEEL ft- 
^(COVER PLATE) 




-ANCHOR TABS 
(FIELD WELDED) 



f?..7T. S a " l/4 GASKET 
ifft ^ £tT» /8 (PLACED AFTER 
DRIVING) 



- ANCHOR TABS 
AND BOLTS 



DETAIL OF PILE 
HEAD W/COVER PLATE 
(ENTIRE ASSEMBLY SET 
AFTER DRIVING) 



-BOOT PLATE 



PILES 2-10 



ITEMS NOT SHOWN: RIBBON CABLE, ACCELEROMETER CABLE, THERMISTOR WIRE. 

SEE INDIVIDUAL DRAWINGS FOR DETAILED DIMENSIONS. ABOVE DIMENSIONS ARE NOMINAL. 

LPC'S, LPC ANCHORS, LPC TUBING, THERMISTORS OMITTEO ON PILES 611 
ACCELEROMETER AND DRIVING STRAIN GAGES OMITTED ON ALL PILES EXCEPT 1,2,4,5. 

* POSITION ON PILES 4 8 5 ** POSITION ON PILES I 82 

FIGURE 1.5. INSTRUMENTATION OF TEST PILES (SCHEMATIC) 
(1 ft = 0.305 m; 1 in = 25.4 mm) 

9 



SITE N 




LEGEND 

# 25' DEPTH SETTLEMENT POINT (DSP) 
AND DEPTH OF ANCHOR 



• SURFACE SETTLEMENT POINT (SSP 

▼ PNEUMATIC PIEZOMETER; DEPTH =19' 

V PNEUMATIC PIEZOMETER; DEPTH = 34' 

V PNEUMATIC PIEZOMETER; DEPTH=50' 

■ REF. BM. MONITOR POINT (SSR) 

NOTE: NUMBER DESIGNATIONS OF PILES AS 
DRIVEN ARE SHOWN ADJACENT TO 
EACH PILE 



SSP5^ 



SEE DETAIL OF PILE ALIGNMENT FOR EXACT 
LOCATIONS OF DSPI AND DSP2 AND OF 
SP4A , SP5A AND SP7A 



FIGURE 1.6. GROUND INSTRUMENTATION PLAN 
(1 ft = 0.305 m; 1 in = 25.4 mm) 

10 



Ground Instrumentation 

Figure 1.6 depicts the layout of the test site and also describes 
the location of ground instrumentation in plan view. Ground 
instrumentation consisted of two principal systems: (1) pneumatic 
piezometers, and (2) Borros-type heave-settlement points. Figure 1.7 
shows the east elevation of the ground instrumentation, indicating the 
depths at which the various instruments were situated. Both systems 
were installed several weeks before the test piles were driven and were 
monitored during test pile installation. The settlement points were 
monitored by means of a microhead level during installation and by dial 
gages suspended from the reference beams during load testing. Exact 
as-built positions of the settlement points nearest the piles are shown in 
Fig. 1.20. 

Loading System and Testing Sequence 

The pile group was loaded vertically by a system of four hydraulic 
jacks hooked to a common pressure manifold and reacting against two 
plate griders. The plate griders were anchored by vertical Dywidag 
bars that were secured to the bases of two concrete anchor caissons 
approximately 105 ft. (32.0 m) below grade. These anchors are 
described in detail in the Interim Repor t. The resulting reaction system, 
consisting of the girders and flexible~bars was essentially articulated, 
rendering it free to translate or to rotate with the pile cap. 

The reference piles were loaded by single jacks reacting against 
beams which were each supported by four H - piles driven to a 
penetration of 25 ft. (7.6 m). Cross beams were placed between the 
reference pile reaction systems to provide a rest for the group reaction 
girders when they were not being loaded. 

All load tests (with one exception, described later) were conducted 
by applying a small increment of compression load (about one-eighth of 
the failure load) each hour. Most instruments were read 5 minutes, 30 
minutes and 55 minutes after each load was applied. Loading was 
monotonic and continued until failure occurred, after which load was 
removed in several decrements (except in the uplift tests, where all 
load was removed in one decrement). The total length of time required 
for each test was on the order of 12 hours. 

The 9-pile group was tested to failure three times: at 20, 82, and 
110 days after the completion of installation, in order to assess the 
effects of set-up. Approximately five days prior to each nine-pile 
group test, the two reference piles were tested simultaneously. The 
first test on Reference Pile No. 11 was a "quick" test in which load 
increments were applied every 2.5 minutes instead of every hour. 

Following completion of the three 9-pile group tests and associated 
reference pile tests, the corner piles (denoted Piles 4,6,8, and 10 on 



11 



27'6"T0 



SSP 5 



2'6" 



SSR4 



EF. 

BM. 



TO 15' 



'6° — J2VJ— 



20 



SSP 



»P3 t 



19* O" 



P 195 



6*0" 



I5'0" 



ie'o" 



vV^N - 



SSP2 



• «SSPI 



elz^ 



5'5'/ 4 "- 



8*3- 
5'Kf- 



4,5,6 2,3,7 8,9,10 

n n n 



ii ii • i 

ii i i it 

■ i i i i , 

i i 



PI93 



PI9<, 



P345 
V 



V 

P192 



i 






fr 



P342f 
PMir 
P344 



H 

V 



("5" 



7?^ 



P194 ▼ 



•-1 5 



-e-v 



DSP I 
{2't 



•OSPi 
(25") 



he'o" 



*—6 ± 



0SP2 
(2'J 



•0SP2 
(25") 



P343 



•OSPI •DSP2 



Ovxw ;kst*J\ 



— .2'0*K-4*0"— • 

SSRzj 



(43') 



(43"> 



P503 



▼ ▼ »DSPI »0SP2 

P50I, P502 (50') (5o'j 

P504 



SSP6 



U 



REF. 

"BM. 



SSP 4 



» TO 15' 
STRATUM 1 

STRATUM 1 1 



STRATUM III 



STRATUM IV 



STRATUM V 



N-S SECTION 



FIGURE 1.7. GROUND INSTRUMENTATION (EAST ELEVATION) 
(1 ft = 0.305 m; 1 in = 25.4 mm) 



STRATUM VI 



12 



Fig. 1.6) were detached from the pile cap and the remaining five piles 
were tested as a subgroup. Immediately following the 5-pile subgroup 
test the center pile (denoted Pile 2) was detached from the cap and the 
remaining four piles were tested as a subgroup. Following the 4-pile 
subgroup test the cap was completely removed, the reaction system 
rearranged, and six of the piles (both reference piles and four group 
piles) were subjected to individual uplift, or tension, tests. The 
chronology of the tests and other significant field events is presented 
in Table 1.1. 

Vertical deflections of each pile in every test were measured by 
two dial gages mounted diametrically opposite each other approximately 
one ft. (0.3 m) off the ground. The dial gages were suspended from 
common reference beams. Backup vertical deflections were also obtained 
by means of a microhead level sighting on scales placed on the pile cap 
and backsighting on a bench mark on a structure outside the zone of 
influence of the piles. 

Twelve dial gages were also mounted at the lower corners of the 
cap to monitor cap rotation and translation in six degrees of freedom 
and to verify cap rigidity. 

In order to minimize errors associated with thermal effects the 
entire test site was covered with a canvas shroud. In order to minimize 
errors associated with moisture content changes in the near- surface 
soils, the testing was carried out in the late fall, winter and early 
spring (wet season at the test site). 

During the first load test on the 9-pile group, tipping of the cap 
toward the north was observed. This is believed to be due partially to 
the fact that the reaction griders were not centered exactly over the 
center of reaction of the piles when the Dywidag bars were vertical. 
Since it was necessary to position the jacks essentially under the webs 
of the girders, the jacks themselves were slightly displaced (perhaps 50 
mm) from the center of reaction. To alleviate this problem on sub- 
sequent , tests the girders and jacks were moved so that the resultant 
jack load would be at the calculated center of pile reaction from the 
first test. This necessitated moving the girders several inches south. 
In order to prevent their translation back to the north in subsequent 
tests, lateral braces were installed on the cross "resting" beams. 
These braces permitted vertical movement but effectively restrained 
horizontal translation of the reaction girders. 

Pile Driving 

General. The reaction piles for the reference test piles (H-piles) 
and reference beam support piles (also H-piles) and the eleven test 
piles were driven in the sequence described in Fig. 1.8. Approximately 
one week was required to install all 23 piles and to retap five of the 



13 



TABLE 1.1. CHRONOLOGY OF MAJOR FIELD EVENTS 



EVENT 


DATE 


Initial Site Investigations 


15 Jan 79 - 


30 Mar 79 


Installation of Anchor Caissons 


27 Aug 79 - 


5 Sep 79 


Installation of Ground Instruments 


6 Sep 79 - 


25 Sep 79 


Installation of H-Pile Anchors 


26 Oct 79 




Predrilling of Pilot Holes 


26 Oct 79 




Installation of Test Piles 


29 Oct 79 - 


1 Nov 79 


Retapping of Test Piles 


2 Nov 79 




Erection of Cap, Reaction and 


5 Nov 79 - 


14 Nov 79 


Reference Frames 






Reference Pile Tests No. 1 


16 Nov 79 




9-Pile Group Test No. 1 


21 Nov 79 




Reference Pile Tests No. 2 


18 Jan 80 




9-Pile Group Test No. 2 


22 Jan 80 




Reference Pile Tests No. 3 


14 Feb 80 




9-Pile Group Test No. 3 


19 Feb 80 




5-Pile Subgroup Test 


26 Feb 80 




4-Pile Subgroup Test 


29 Feb 80 




Uplift Test: Pile 1 


19 Mar 80 




Uplift Test: Pile 11 


21 Mar 80 




Uplift Test; Pile 2 


27 Mar 80 




Uplift Test: Pile 9 


28 Mar 80 




Uplift Test: Pile 5 


1 Apr 80 




Uplift Test: Pile 4 


3 Apr 80 




Final Soil Soundings and 


8 Apr 80 - 


18 Apr 80 


Site Closure 








i 





14 



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m. 

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<> 



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17 



R6 



8'- 6" 



ANCHOR LOCATION 
(TYP. BOTH SIDES) 




2'-9" 



II 



2'- 9" 



H 



o — & — & 



8'-6" 



e — ^5—^ 



14 



n 

e — =0— d© 



NOTE : 

NUMBER IN BOX INDICATES 
DRIVING SEQUENCE. DISTANCES 
ARE NOMINAL. R PILES ARE 12 H53; 
RB PILES ARE 8 H36 TEST PILES 
ARE 10.75 IN. <£X 0.365 WALL CLOSED- 
ENDED PILES 



R3 



H- 



i 7" 

3 L 5— 
~I6 



i 



m 



R4 



6 



15 



R7 



& 






RB3 



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CM 



0> 
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0> 
I 

"co 



CO 

_l 
CM 



CO 
"CM 



Moo 11 

■fit 



FIGURE 1.8. APPROXIMATE LOCATIONS OF ALL TEST AND REACTION PILES AND 
SEQUENCE OF DRIVING (1 ft = 0.305 m; 1 in = 25.4 nun) 



15 



test piles. Each of the test piles was driven in an 8-inch (200 mm) 
diameter by 10 ft. (3.1 m) deep pilot hole in order to ensure proper 
positioning and vertical alignment. All of the pilot holes were drilled at 
one time, 3 days before the first test pile was driven. The site was 
innundated with a heavy thunderstorm on the night of October 30, 
1979, following the driving of Pile No. 9 (12th pile in the sequence). 
This left the pilot holes for the remaining test piles containing some 
free water which the construction crew attempted to remove. This 
effort was only partially sucessful. The presence of free water in the 
pilot hole for Pile 11 may explain the anomalous behavior experienced by 
this pile during the first load test, described later. 

All 23 piles were driven by a Raymond IS hammer, which is a 
single acting steam hammer with a rated energy of 19,500 ft-lbs. 
(26,500 m-N) per blow. The steel driving cap for the test piles fit 
over the outside ,of the piles. Cushioning, in the form of a 16.5-in. 
(404 mm) thick stack of alternating pads of aluminum and micarta was 
employed between the hammer and the driving cap. There was no 
cushion between the driving cap and the pile heads. The hammer was 
supported on hanging leads without a spotter beam. Driving records 
for all 23 piles are presented in Table 1.2. Note that Piles 1,2,4,8, 
and 11 were initially driven slightly short of their intended penetration 
and then retapped to final penetration at varying times after initial 
driving. These "retaps" were intended to serve as a qualitative guide 
to short-term set-up. 

Quantitative data on actual energy delivered to the pile heads and 
on short-term set-up were acquired by means of the "Goble Pile Driving 
Analyzer" by others during installation. The reader is referred to the 
FHWA report "Dynamic Pile Driving Measurements for University of 
Houston Pile Group Study" by A.R. Dover, G.E. Locke, and J.M.E. 
Audibert, dated December, 1979, for that imformation. 

Some comments on the above report appear appropriate here. A 
review of that report reveals that actual energy accepted by the pile 
heads varied from about 4,000 ft-lbs (5,400 m-N) per blow for Pile 9 to 
about 8,000 ft-lbs (10,900 m-N) per blow for Pile 11. These values had 
very little dependence on tip penetration. Short-term set-up, based on 
retap data and Analyzer results requires the assumption of a damping 
factor, denoted J. Prior experience by the authors of the above- 
referenced report suggests that the appropraite J-factor is that which 
applies to the soil type near the pile tip. In this case (sandy clay) J 
should be in the range of 0.45 - 0.70. Analysis of the predicted rate 
of set-up over a period of four days (lapse period for the retaps) with 
J = 0.6, at the approximate midpoint of the time range (Fig. 13 of the 
above-referenced report) suggests that the static pile capacity during 
driving was likely between 50 and 100 kips (220 and 450 kN) and 
that the static capacity approximately four days after driving was 150 
to 200 kips (670 to 890 kN) for Piles 1 and 2. The implications of 
these data, if a constant J-factor is realistic, are that almost all of the 



16 



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17 



set-up of both the center group pile (No. 2) and the isolated reference 
pile (No. 1) occurred within four days of driving, since the capacities 
of Piles 1 and 2 were in the range of 150 to 170 kips (670 to 760 kN) at 
the time of Test No. 1, approximately three weeks after driving. This 
result appears to be consistent with the rapid rate of pore water 
pressure dissipation observed at this test site following driving. Pore 
pressure effects are described in detail later. 

Measurement of Compression Wave Attenuation . Representative 

traces of dynamic force vs. time during driving are presented in Figs. 
1.9 and 1.10. Figure 1.9 shows force-time traces at the tops, near the 
middles, and near the bottoms of Pile 2 and Pile 4 for one hammer blow 
at essentially full penetration during initial driving . Exact locations of 
transducers are noted on the figure. Two transducers were placed at 
opposite ends of a diameter at the top locations to allow the bending 
effect produced by an uneven blow to be accounted for (by averaging 
the traces for the two gages). At the lower levels only one transducer 
was placed because it was assumed that effects of eccentricity of the 
stress wave would be damped out by the soil by the time the wave 
reached those locations. The average force-time trace for the top 
transducers, as well as all individual force-time traces are shown with 
respect to a common time base for each pile. Two facts are evident in 
these traces: (1) very little force attenuation occurred over the first 
21 ft. (6.4 m) of embedment, and (2) considerable force was trans- 
mitted to the vicinity of the pile tips. These observations imply that 
virtually all of the soil resistance developed during driving was in the 
form of side shear over the bottom half of the pile and of tip 
resistance. This inference is in general agreement with the wave 
return measurements made at the pile heads by Dover, et al, using the 
Goble Pile Driving Analyzer. 

For purposes of futher analysis, it may be assumed that Pile 2 
behaved as an "isolated" pile during driving, since it was the first pile 
driven in the group, and that Pile 4 behaved as a "group" pile, since 
it was one of the last piles driven in the group. 

Similar driving traces are shown in Fig. 1.10 for Piles 1 and 5. 
Pile 5 data are shown for a partial penetration of 28 ft. (8.5 m). 
Beyond that penetration one of the top gages and the botton gage 
yielded a flat response, probably due to a broken wire or connection. 
Pile 5 is also a "group" pile, as it was the last pile driven. The partial 
pentration data for Pile 5 indicated a significant reduction in transmitted 
force in the upper 22 ft. (6.7 m) of soil, unlike the essentially non- 
existent reduction in the upper 21 ft. (6.4 m) of soil in Piles 2 and 4 
at full penetration. The same general effect was observable for Piles 2 
and 4 for partial penetrations. This suggests that effects of driving 
may have degraded the contact between the pile wall and the soil in the 
upper approximately 20 ft. (6.1 m) as the piles were being driven from 
about 28 ft. (8.5 m) of penetration to 43 ft. (13.1 m) of penetration. 
Subsequent load testing, however, indicated relatively high load transfer 



18 



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in the upper 20 ft. (6.1 m), suggesting that the soil in this zone may 
have developed a very small annular space with respect to the pile wall 
during driving after which the soil rapidly swelled laterally into contact 
with the pile wall. 

The data for Pile No. 1 are incomplete. One of the top gages 
saturated electrically because of an improper amplifier setting. The 
bottom gage yielded a flat response. This gage was later found to be 
faulty. The scheme of establishment of dynamic gages with respect to 
bending was identical on Piles 1 and 5 to the scheme described for Piles 
2 and 4. 

Wave Equation Analyses . Force measurements made during pile 
installation at three locations on instrumented piles indicate some 
attenuation or reduction in pile shaft force between gage stations two 
(GA 2) and three (GA 3) (Fig. 1.11c) for readings made just prior to 
final penetration. Wave equation modeling was attempted for Piles 2 and 
4. Various values of Smith side damping were examined to determine 
what comments could be made relative to appropriate values for use in 
analysis of a single pile (Pile 2) versus a group pile (Pile 4). 

Recorded driving resistance for Pile 2 was 1.83 blows per inch 
(0.072 blows /mm), and the pile force was observed to drop from 300 
kips (1330 kN) at GA 2 to 60 kips (267 kN) at GA 3. Recorded driving 
resistance for Pile 4 was 1.5 blows per inch (0.059 blows/mm), and the 
pile force was observed to drop from 350 kips (1560 kN) at GA 2 to 200 
kips (890 kN) at GA 3. Evaluation of the force- time records indicate 
that approximately 40 per cent of the static pile capacity developed 
during driving occurred at the tip and the remaining capacity occurred 
in side friction over approximately the bottom one half of the penetra- 
tion length. These proportions were used in the wave equation study. 
The wave equation program used is the program developed by E.A.L. 
Smith while he was Chief Mechanical Engineer for Raymond International. 

The results of the study are indicated on Figs. 1.11 and 1.12 and 
in Table 1.3. The best match of measured versus calculated pile force 
was obtained for Pile 4 using a uniform friction distribution and a 
J-side of 0.4. Comparision of measured versus calcualted pile force 
(Fig. 1.12a) indicates that for Pile 4 the measured forces were 
approached asymptotically by the calculated forces as J-side was 
increased, indicating that reasonably larger values of J-side would not 
further enhance the comparison. Similarly, this figure also indicates 
that the measured force at GA 3 for Pile 2 would never be reached 
using the R distribution under study. 

Conclusions which can be drawn from this comparison of measured 
and calculated forces are (1) for the distribution under study a J-side 
of 0.4 appears appropriate for group Pile 4, and (2) either a different 



21 



MEASURED P4GA2 



950 



300 



CALCULATED 
-P4GA2 
P4GA3 — 



a. 



tu 



250 



200 



150 



100 



50 



v> 






100 



75 



50 



25 ■ 




-CALCULATED P2 6A3 



MEASURED P2 6A 3 



J- 



-L 



0.1 



0.4 



0.2 0.3 

SIDE DAMPING, J, 

(o) MEASURED AND CALCULATED 
PILE FORCE 



PILE 2 ® 183 BPI 




PILE 4 (£ 1.5 BPI 



-L 



X 



0.1 0.2 0.3 

SIDE DAMPING, J, 

(b) DYNAMIC RESISTANCE vt. J. 



0.4 



RAYMOND 
IS HAMMER 
WRAM:6500LB 
M/AL CAPBLOCK 
K 

FOLLOWER 
W=I600 LB 



3. 74 k 10* L % 









H 


hi' 




4 






5 




GAI . 


6 




(PILES 284) 


7 


W&// 




8 


I 




9 


A 






>C 




10 






II 

"IF 




J3_ 
14 
15 


1 




il 




GA 2 


17 


29' 


(PILES 2 8"4T 


16 
19 
20 

1L 

22 
23 
24 


— 31' 


GA3 




43' 


(PILE 4P 


25 
26 
27 


45' 


GA3_ 


28 


49 


(PILE 2)1 




51.25* 



(c) WAVE EQUATION 
MODEL 
PILES 2 8 4 



FIGURE 1.11. MEASURED AND CALCULATED PILE FORCES AND WAVE 
EQUATION MODEL (1- k = 4.45 kN; 1 ton = 8.9 kN; 1 ft 
= 0.305 m; 1 lb = 4.45 N; 1 in = 25.4 mm} 



22 



PILE 4 



MEASURED FORCE 

WAVE EQUATION CALCULATED 

FORCES 

WAVE EQUATION SEGMENT 

VELOCITY ■ •— 




V TIME (MILLISECONDS) 



MIDDLE GAGE @. 29' 




■i i i /i i i % ■! i i iini 

V 7$ v V 



TIME (MILLISECONDS) 



300 



BOTTOM GAGE <S 44' 




TIME (MILUSECONDS) 



30 



FIGURE 1.12. MEASURED AND COMPUTED FORCE-TIME RELATIONSHIPS 
FOR PILE 4 (lk= 4.45 kN; 1 ft = 0.505 m) 

23 



TABLE 1.3. COMPUTED PEAK PILE FORCES DURING DRIVING AS 
FUNCTION OF SIDE DAMPING, J (1 k = 4.45 kN; 
1 t = 8.9 kN; 1 ft = 0.305 m; 1 in = 25.4 mmj 
1 ft-lb = 1.36 m-N; 1 BPI = 1 blow / 25.4 mm) 



Pile 
Pile 


4 Measurec 
2 Measurec 


1 
I 




MAXIMUM PILE FORCE, KIPS 






@6'-6" 
300 K 
300 K 


@29'3" 
350 K 
300 K 


@ 44 ' 
200 K 


@ 50'-3" 
60 K 


40% 


Point, Ti 


iangular Fric 


tion Distribution Lower 


23' 




Pile 


*u 




Js 


GA 1 


GA 2 


GA 3 


GA 3 


4 
2 


75 T 
88 T 




0.05 
0.05 


299.3 
299.3 


318.4 
321.7 


282.2 


320.2 


4 
2 


58 T 

70 T 




0.20 
0.20 


299.3 
299.3 


328.4 
334.2 


269.1 


180.2 


4 
2 


52 T 
63 T 




0.30 
0.30 


299.3 
299.3 


333.8 
340.8 


265.6 


165.1 


4 
2 


46 T 
57 T 




0.40 
0.40 


299.3 
299.3 


337.3 
345.6 


262.1 


153.8 


40% 


Point, Uniform Frictio 


n Distribution Lower 23' 






Pile 


Ru 




Js 


GA 1 


GA 2 


GA 3 


GA 3 


4 
2 


75 T 
89 T 




0.05 
0.05 


299.3 
299.3 


322.9 
327.2 


251.0 


209.5 


4 
2 


59 T 
70 T 




0.20 
0.20 


299.3 
299.3 


335.0 
340.8 


222.1 


162.7 


4 
2 


51 T 
62 T 




0.30 
0.30 


299.3 
299.3 


339.4 
346.6 


210.0 


144.8 


4 
2 


46 T 
56 T 




0.40 
0.40 


299.3 
299.3 


343.3 
351.0 


201. * 


132.0 


NOTES : 1 . 
2. 

3. 


Pile 4 driven 
Hammer impact 
delivered by 
Side quake & 
damping = 0.1 


to 1.5 BPI, Pile 2 to 1 
velocity = 12.81 ft/sec 

hammer = 16,575 ft-lbs. 

Point quake = 0.10, and 

5 for all runs. 


.83 BPI 
or energy 

Point 





24 



R distribution existed at Pile 2 or significantly higher damping 
occurred during installation of this pile. Examination of the R curves 
of Figure 1.11b indicates that for the distribution studied, Piles 2 and 
4 had a minimum capacity at the time of driving of approximately 57 and 
46 tons, respectively. These values also seemed to be approaching a 
limit as J-side was increased. Damping associated with transverse 
vibration of the pile could have been active during driving, but such 
damping is not modeled in the one-dimensional wave equation. 

The maximum pile forces reported by the Pile Analyzer equipment 
at a point approximately three feet (1 m) above GA 1 for Piles 2 and 4 
at approximately the same penetration are 157 and 256 kips (699 and 
1139 kN), respectively, while measurements made for this study 
consistently indicate 300 kips (1335 kN). Similarly, maximum energy 
transferred to the piles as calculated by the integral of force times 
velocity over time for Piles 2 and 4 were, respectively, 4400 and 6500 
foot-pounds (6000 and 8840 m-N). Matching of wave equation analysis 
to the measured 300 kips (1335 kN) pile head force indicates an 
effective required hammer energy of 16,575 foot-pounds (22,540 m-N). 
Calculation of energy delivered to the pile, as opposed to energy 
accepted by the pile (and presumably dissipated into the soil), using 
the measured velocity for Piles 2 and 4 from the Pile Analyzer and the 
hammer ram weight in the expression E = \ M V 2 yields a delivered 
energy of 7640 and 10,296 foot-pounds (10,390 and 14,000 m-N), 
respectively, for Piles 2 and 4. These values seem low, but they may 
offer a more realistic basis for hammer evaluation than does the energy 
transferred to the pile given directly by the Analyzer. 

Soil Displacements During Driving 

The surface and depth settlement points ("SSP" and "DSP," 
respectively) were monitored during test pile installation by means of a 
microhead level. This level was realistically capable of resolving 
movements of 0.05 in. (1 mm) or greater. The results of this monitoring 
are summarized in Fig. 1.13, in which the progression of soil movement 
along a north-south line is shown at various stages of driving. The 
maximum heave of the soil surface was about 1 in. (25 mm) near the 
outer perimeter of the group, reducing to about 0.1 in. (2.5 mm) at a 
distance of 28 ft. (8.5 m) from the center of the group. The reference 
piles were driven along this north-south line, and they appear to have 
magnified the heave in the soil some distance from the group. 

Below-surface vertical movements were generally small except at a 
depth of 25 ft. (7.6 m), where about 1 in. (25 mm) of settlement was 
noted in the settlement point nearest Pile 9 (DSP1). This phenomenon is 
probably a result of one of the spread anchors for the point being in 
the zone of shear drag for this pile. Farther from the group perimeter, 
at DSP2, slight heaving of the soil was observed. The observations 
imply a vertical extension of the soil near the perimeter of the group in 
the upper 25 ft. (7.6 m). Very little net movement was observed at 



25 



GROUP G_ 



ORDER OF DRIVING 
1,2,3,10,9,8,7,6,11,4,5 



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cr £■ 

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PILE 



LEGEND 

• — AFTER DRIVING PILE I 

AFTER DRIVING PILE 3 

AFTER DRIVING PILE 9 

AFTER DRIVING PILE 6 

AFTER DRIVING PILE II 

DRIVING COMPLETED 



PILE 



PILE I 



4,5,6 10,9,8 
\ PILE 



PILE 1 1 




+F 



i 



I 



dfe r .Ar - 



5' 5" 



3'5"3'0'.'*|» 



■Ft 

i'ii" 

■ i ■ i. ii i .ii 



SSP3 SSP2 

SSPI 



4' I" 



DSPI 
DSP2 



REFERENCE 

SURFACE 

(2 FT. DEPTH) 



NOTE : RESOLUTION OF 
MOVEMENTS WAS~0.05IN. 
NO DETECTIBLE MOVEMENTS 
AT SSP5 



6'2" 



I0'0" 



(25 FT. DEPTH) 



(43 FT. DEPTH) 

(BOTTOM OF 
PILES) 



(50 FT. DEPTH) 



x 2 9 



SSP6 



SSP4 



N - S PROFILE LOOKING WEST 



FIGURE 1.13. 



OBSERVED SOIL MOVEMENTS DURING PILE INSTALLATION 
(1 ft = 0.305 m; 1 in = 25.4 mm) 



26 



DSP1 and DSP2 at a depth of 43 ft. (13.1 m), although it should be 
recalled that these points were positioned outside the perimeter of the 
group. Slightly greater movements (settlements) were observed at the 
50 ft. (15.3 m) depth. 

The observed surface heave accounted for approxmately 30 per 
cent of the theoretical volume of desplaced soil from the 11 test piles, 
assuming the displacements were symmetric; that is, displacement 
measured along the north-south line would apply to any other section 
through the group. 

Pore Water Pressures 

Figures 1.14-1.17 present the pore water pressure history, as 
measured by the pnuematic ground and pile piezometers from prior to 
the time of installation through the first load test. The free-field 
response of 11 of the 14 ground piezometers for the 30-day period prior 
to pile installation is shown in Fig. 1.14. All but 3 of these 
piezometers reached essentially a steady state condition prior to pile 
installation. The three that did not reach steady state drifted toward 
higher indicated pressure readings rather than lower readings. One of 
these piezometers, P503, eventually returned to a reasonable reading. 
The other two (P343 and P504) continued to drift upwards in reading 
throughout the test program and are considered by the authors to have 
yielded unrepresentative values of pore pressure. Two other 
piezometers, P194 and P341, gave zero pore pressure readings, (flat 
response) throughout the test program. Thus, it appears that 10 of 
the 14 ground piezometers yielded reasonable results through all or part 
of the testing program. The performance of both the pile and ground 
piezometers is discussed further in Appendix E. 

The pile piezometers lost saturation upon driving. Therefore, 
pore pressure readings against the pile faces immediately after driving 
could not be made. The pile piezometers were resaturated during and 
immediately after the period of pile installation through special tubing 
that had been installed for that purpose. The first valid reading on 
the various pile piezometers after resaturation is denoted by the 
character "S" on Figs. 1.15-1.17. These figures do not show the 
pressure- time relationships for the 9 ft. (2.7 m) and 41 ft. (12.5 m) 
pile piezometers. The 9 ft. (2.7 m) piezometers registered essentially 
zero response and the 41 ft. (12.5 m) piezometers responded in a 
manner similar to the 34 ft. (10.4 m) piezometers. Numerical values for 
these levels are given in a different form later in this report. 

Several observations from Figs. 1.15-1.17 are noted. First, the 
ground piezometers yielded free-field pore pressures (before installation 
of piles) that could be approximated with a piezometric surface at a 
depth of about 7 ft. (2.1 m). The average hydrostatic piezometer 
reading before pile installation at each level was as follows: 



27 




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31 



Piezometer Depth Depth of Hydrostatic 

Head Level Below Site Grade 
9 ft (2.7 m) 8.0 it~j2A~mT 

19 ft (5.8 m) 7.5 ft (2.3 m) 

34 ft (10.4 m) -4.8 ft (-1.5 m) 

50 ft (15.3 m) 7.3 ft ( 2.2 m) 

The negative value at the third level indicates that the appropriate 
hydraulic head line was above grade. This value is believed to be 
anomalous and unrepresentative due to one high reading. Some scatter 
in the hydrostatic pressure values at a given level can be observed. 
This scatter is believed to be the result of local variations in soil 
properties and their effect on piezometer response. 

Second, the action of driving the "non-displacement" H-type 
reaction piles elevated the pore water pressures, especially at P345, 
near Pile 1. These pressures were largely dissipated three days after 
the H-piles were driven and prior to driving the test piles. 

Third, pore pressure response to test pile installation in the soil 
mass showed that in general rather large excess pore pressures were 
generated by pile installation and that these excess presures dissipated 
rapidly. The functional ground piezometer at the 19 ft. (5.8 m) level 
within the pile group (P192) experienced a peak pressure of 44 psi (303 
kN/m 2 ), 39 psi (269 kN/m 2 ) of which was excess. The peak developed 
excess pore pressure may be expressed as a ratio to the existing 
vertical in-situ effective stress. This ratio, hereafter called the pore 
pressure ratio, was 3.3 for P192 immediately after the installation of 
Pile 2. The pore pressure at P192 dissipated rapidly but responded 
sharply again when Piles 9 and 7 (near the piezometer) were driven. 
These later responses were not as strong as the response associated 
with the installation of Pile 2. The peak readings, expressed as pore 
pressure ratios, were 2.5 upon installing Pile 9 and 1.9 upon driving 
Pile 7. The corresponding piezometer adjacent to Pile 1 (P195) 
responded similarly to the driving of Pile 1, with a peak pore pressure 
ratio of 2.0. The effects of driving other piles were less pronounced 
at P195 than at P192, which was inside the group. P193, which was 
situated between the group and Pile 1 responded less intensely to the 
installation of the test piles, exhibiting a gradual buildup of pore 
pressure during installation of the group rather than sharp response to 
the installation of any single pile. The maxium pore pressure ratio 
was 0.6 and occurred after the completion of the retaps. 

At the 34 ft. (10.4 m) level, somewhat less consistent behavior 
occurred. The two functional ground piezometers within the group 
responded somewhat differently. P344, between Piles 2 and 3 did not 
respond when Pile 2 was driven but did respond to the driving of Pile 
3. The pore pressure ratio was 0.6 after Pile 3 was driven. Pore 
pressures at this location continued to increase until after Pile 6 was 
driven, at which time the pore pressure ratio was 1.2. Thereafter, 



32 



rapid dissipation occurred until the first group test, at which time 
the excess pressure was about 3 psi (21 kN/m 2 ). P342, between Piles 
3 and 4, responded only slightly to the installation of Piles 3 and 4. 
Instead, a general increase in pore pressure occurred during 
installation of the test piles. A maximum pore pressure ratio of 0.6 was 
achieved after the final retap. However, this piezometer did not exhibit 
any dissipation of the excess pore pressure after the retaps, such 
that about 12 psi (83 kN/m 2 ) of excess pore pressure existed at the 
time of the first group test. This is inconsistent with the behavior of 
other ground piezometers and is thought to be associated with a soil 
anomaly in the immediate vicinity of the piezometer. P345, adjacent to 
Pile 1, also failed to respond appreciably to the installation of Pile 1 
but did exhibit a slight pore pressure buildup during installation of 
group piles, followed by rapid dissipation back to the hydrostatic level. 

Pore pressure changes in the soil beneath the level of the pile tips 
were minor. The peak pore pressure ratio at P502, situated essentially 
beneath the center of the group at a depth of 50 ft. (15.3 m), was 0.2 
and occurred just after driving Pile 6. Reestablishment of hydrostatic 
pressures occurred in the functional piezometers at the 50 ft. (15.3 m) 
level within 1 to 4 days after completion of driving. 

Several of the ground piezometers were monitored while piles were 
in the process of being driven. Piezometers near driven piles that 
responded during driving of the piles did so as the pile tips approached 
the depths of the piezometers. After the tips passed, very little further 
pore pressure changed occurred. 

Fourth, the piezometric changes in the soil mass during load 
testing were very minor compared to those created by pile installation. 
(While the absolute values of the drifting piezometers, P343, P 503, and 
P 504, may be incorrect, the registered changes in pore pressures 
during the course of a load test are thought to be valid.) This fact 
suggests that the soil was behaving in essentially a drained manner. 

Fifth, the response of the pile piezometers, once they had been 
resaturated, exhibited a trend similar to the ground piezometers in that 
pore pressures, both on the reference and group piles, decreased 
rapidly after driving and that excess pore pressures generated during 
load testing were very small. 

The rapid pore pressure dissipation observed at this site is 
believed to be due to the high hydraulic diffusivity of the soil produced 
by fissure and slickenside planes in the Beaumont soil, by apparently 
continuous sand partings in the underliying Montgomery soil, and by 
the apparent inability of the pile to induce lateral pressures sufficient 
to transform the soils near the surface to a state of normal consolidation. 

Total pressure cells on the piles were also read after pile 
installation. A rapid temperature drop on the cells during the first few 



33 



hours after driving rendered interpretation of the data for this event 
questionable. Further discussion of total pressure will be deferred 
until Chapter 2; however, for completness with respect to pile 
installation, Fig. 1.18 is shown here, which depicts the average 
measured earth pressure coefficients against the reference and group 
pile walls compared with the in-situ values. The low value at 19 ft. 
(5.8 m) in the reference pile is due to an unrepresentatively low total 
pressure reading. 

Figure 1.19 depicts section lines through various soil and pile 
pressure cells. These sections will be referenced later in this report 
when detailed descriptions of pore and total pressure changes are 
discussed. Meanwhile, Fig. 1.20 shows three of the sections along 
with a spatial comparison of measured pore pressure distribtuions one 
day before driving began and one day after driving was concluded. 

Assessment of Soil Disturbance 

At the conclusion of the tests that will be described in Chapter 2, 
approximately 5 months after the piles were installed, a final series of 
static cone soundings was made. The locations of these soundings, as 
well as the detailed results, are described in Appendix C. No evidence 
of shear strength reduction in the soil within the pile group could be 
found at that time. The penetrometer was not capable, however, of 
making precise measurements of shear strength in the zone immediately 
adjacent to the pile walls. 

As-Driven Locations of Piles 

After the piles had been driven and the cap secured, the locations 
of the heads of all of the group piles were located relative to a point of 
reference on the pile cap by survey techniques. The alignment of piles 
below the pile tops was measured by using a sensitive miniature 
electrical inclinometer, which was run down the inclinometer tubes 
(affixied to the interior of the piles) on two perpendicular tracks. The 
results of these measurements are shown on Fig. 1.21. That figure 
also shows the exact position of certain of the ground settlement points 
relative to the piles and of the loading jacks at the time of testing. 

Note that none of the piles was exactly vertical, although only 
minor bending was observed . The average batter was about 1.5 per 
cent of the pile length, or slightly less then the inside diameters of the 
piles. The highest batter was in Pile 8 (18 in. (0.46 m)). There was 
a slight preference for a batter to the northeast or southwest, which 
was generally perpendicular to the plane containing the driving leads 
and the boom of the crane carrying the leads. 

Table 1.4 gives numerical values for the locations of the centers of 
the pile heads and jack bases at the top of the cap relative to the 
geometric center of the top of the cap. 



34 



LATERAL EARTH PRESSURE COEFFICIENT 



5- 



10- 



15 



I- 
I 

h- 
Q. 
UJ 
Q 



20 



25 



30- 



35- 



40- 



(a) 



I 



2 

T 



3 

~r 




MEASURED 
AGAINST PILES 
(AV6. OF GROUP) 



MEASURED AGAINST 
REFERENCE PILE NO. I 



43 L 

FIGURE 1.18. MEASURED EFFECTIVE EARTH PRESSURE COEFFICIENTS AGAINST 
PILES FOUR DAYS AFTER INSTALLATION (1 ft = 0.305 m) 

35 



9- and 19-FOOT LEVELS 



34 -FOOT LEVEL 



o 



TEST PILE (jQ) 
DESIGNATION^ 



PI94 



® T ® 
PI92 




WEST 
ANCHOR 



(p— <|) ® 



.4) © 

03 PI9I 



© 



GROUND 
-PIEZOMETER 
DESIGNATION 



/PI9S 





P343 



P34l\T P344 
^P342\ 

Cw (5) 



© 



P345 



41-FOOT LEVEL 




3) ® ® 



(4) © 



P50 



50- FOOT LEVEL 

© ® (D 




^^ 



FIGURE 1.19. PORE PRESSURE PROFILE SECTION LINES (1 ft = 0.305 m) 

36 



50-1 



PI95 



PI93 



PI9I 
SECTION C-C 



PI92 



PI94 



<6b 
P343 




SECTION K-K 



LEGEND 

BEFORE DRIVING (10/ 25/ 79) 
I DAY AFTER DRIVING (1 1/ 2/79) 



FIGURE 1.20. GROUND PORE PRESSURE PROFILES: BEFORE AND AT CONCLUSION 
OF DRIVING (1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ) 



37 



SITE N 



+ 2' (DEPTH) 
43' 25' 50' 



DSP2 



EDGE OF CAP 




SCALE: 



10" 



20" 



30' 



40" 



NOTE: INSIDE DIAMETERS OF PILES ARE 
SHOWN. C_ POINTS SHOWN AT 
10 FT. INTERVALS. 

NOTE: SSP4A, 5A.AND 7A WERE NOT 

POSITIONED UNTIL AFTER FIRST 
GROUP TEST. 



FIGURE 1.21. AS -CONSTRUCTED LOCATIONS AND ALIGNMENTS OF GROUP PILES 
(1 ft = 0.305 m; 1 in = 25.4 mm) 

38 



TABLE 1.4. PILE HEAD AND JACK COORDINATES FOR 
9-PILE TEST NO. 1 (1 in. = 25.4mm) 





\ PILE NO. 




X (in.) 


Y (in.) 


Z (in.) 


2 
3 
4 
5 
6 
7 
8 
i 9 
10 


- 0.8 
-32.0 
-33.0 

- 0.5 
+33.0 
+32.0 
+ 31.8 

- 2.0 
-30.9 













- 1.8 

- 1.8 
+30.0 
+29.8 
+31.0 

- 2.0 
-33.5 
-33.0 
-30.4 


JACK NO. 1 
JACK NO. 2 
JACK NO. 3 
JACK NO. 4 


-16.6 
-16.4 
+ 16.4 
+ 16.8 

1 








+ 15.0 
-18.4 
-18.2 
+14.0 



NOTES: JACKS 2, 3, 4 MOVED 3.5 IN. SOUTH (+Z) FOR TEST 2. JACK 1 
MOVED 3.0 IN. SOUTH (+Z) FOR TEST 2. ALL JACKS MOVED 1.5 
NORTH (-Z) FROM TEST 2 POSITION FOR TEST 3. 





■>• *x 



SOUTH ELEVATION 



39 



Finally, the exact locations and alignments of the reference piles 
are shown graphically in Fig. 1.22. 

Residual Loads Developed in Piles Due to Driving 

Zero readings were made on the strain gage circuits when the piles 
were in the calibration bed in an unstrained state and were checked 
again in the field just prior to driving each pile. Once each pile was 
driven a new set of strain circuit readings was made on each of the 
piles in the ground in an attempt to determine patterns of development 
of residual loads in the group piles compared to those in the reference 
piles. Readings taken during the installation process were largely 
unusable, due to instabilities in the data acquisition system, as 
explained in Appendix E. Stable readings were obtained, however, 
prior to the first load tests. These readings are shown graphically for 
the reference piles in Fig. 1.23. The largest residual loads were 
developed in Pile 1. Pile 11 exhibited low residual loads, perhaps 
because the presence of free water in the pilot hole for Pile 11 caused 
greater lubrication of that pile. The observed residual load pattern 
requires that negative side resistance (directed downward on the pile) 
be present above a depth of 25 ft. (7.6 m) and that positive side 
resistance exist below that depth. The average compressive residual 
tip load due to driving the reference piles was about 7 kips (31.2 kN). 

The residual load pattern in the group piles was extremely complex. 
No pattern with respect to geometric position could be positively 
identified, therefore, the authors have chosen to average all of the 
residual load readings and thereby consider a typical group pile. The 
average residual load in the group piles and the measured upper and 
lower bounds are shown in Fig. 1.24. Some of the extreme readings 
may be the result of zero shifts in several of the circuits. Since the 
circuits were wired with random polarity among the piles, errors of this 
type tend to be cancelled when averages are taken. 

The residual loads thus obtained in the group piles were generally 
lower than those in the reference piles, as demonstrated in Fig. 1.25. 
This is an expected phenomenon. The pattern of load distribution 
along the typical group pile was somewhat different from the pattern in 
the reference piles. Positive side resistance was encountered down to a 
depth of 15 ft. (4.6 m), presumably due to the load imparted by the 
cap (57 kips (254 kN) to 9 piles). Below that depth, side resistance 
was negative to a depth of 35 ft. (10.7 m) after which it again became 
positive. The residual tip load in the typical group pile was 3 kips 
(13.4 kN). 

The effect of residual loads on the development of load transfer 
will be considered further in Chapter 3. Due to electrical drift it was 
not possible to maintain predrive zeros beyond the first load test. 
Specific reasons for this problem are discussed in Appendix E. 



40 




A SITE 

r 



V 3 FT. ABOVE 
PILE TIP 



<3CAUE. 
h 10" H 
(PILE DEVIATIONS) 



NOTE : INSIDE DIAMETERS 

OF PILES ARE SHOWN. 
<£ POINTS ARE SHOWN 

AT 10 FT. INTERVALS 
FROM TOP. LAST POINT 
IS 43 FT. BELOW TOP 

NOTE ; LOCATIONS OF 

PILES RELATIVE TO 
GROUP AND PILE 
DEVIATIONS ARE TO 
DIFFERENT SCALES 



FIGURE 1.22. AS -CONSTRUCTED LOCATIONS AND ALIGNMENTS OF REFERENCE PILES 

(1 ft = 0.305 m; 1 in = 25.4 mm) 

41 



RESIDUAL LOAD CK) 
10 20 




UPPER BOUND C+D 
CPILE ID 



/ 31' LEVEL INOPERATIVE 
/ ON PILE I 



FIGURE 1.23, 



RESIDUAL LOADS IN REFERENCE PILES AFTER INSTALLATION 
(1 ft = 0.305 m; 1 k = 4.45 kN) 

42 



RESIDUAL LOAD CJO 
■10 10 20 




NOTE: BOUNDS 
NOT DEVELOPED 
BY ANY SPECIFIC 
PILE 



FIGURE 1.24. 



RESIDUAL LOADS IN TYPICAL GROUP PILE AFTER INSTALLATION 
(1 ft = 0.305 m; 1 k = 4.45 kN) 



43 



a 
u 
o 



AVERAGE RESIDUAL LOAD CKD 
10 20 

nr 



INCLUDES EFFECTS 
OF CAP AND LOAD 
SYSTEM WEIGHT 



ESTIMATED FOR 
GROUP PILES HAD 
PILE CAP BEEN 
ABSENT 




PRE-DRILL 
DEPTH 



i 



AVG. OF REFERENCE 
PILES BEFORE TEST I 



FIGURE 1.25. COMPARISON OF RESIDUAL LOADS (PER PILE) IN REFERENCE AND 
GROUP PILES (1 ft = 0.305 m; 1 k = 4.45 kN) 



44 



Chapter 2 . Pile and Soil Performance Under Load 



General 



This chapter contains partial test results for the seventeen static 
load tests conducted as outlined in Table 1.1. Test results specifically 
pertaining to load transfer are deferred to Chapter 3. The various 
interpretations shown here in graphical and tabular form are based 
upon readings made 5 minutes after each increment of load was applied, 
except for the first test on Pile 11, where 30 second readings were 
used, and where otherwise noted. All reported settlements are as-read 
values that are uncorrected (except where noted) for reference system 
movements, 'and all loads are the loads indicated by the second highest 
level of strain gages (or sum thereof in group tests) on the piles, as 
verified by load cell and jack pressure measurements. 

Load- Settlement Behavior 

Reference Piles . Three tests were conducted on each reference 
pile, preceding by four to five days each 9-pile group test. The 
results of reference pile tests which immediately preceded a given group 
test were used to assess settlement ratio and efficiency (defined later) 
for that test. Separate reference pile tests were not conducted in 
association with the subgroup tests; therefore, the set of reference pile 
tests conducted in conjunction with the third and final 9-pile test was 
also used as a baseline for the subgroup tests. Figure 2.1 depicts the 
load-settlement behavior of Piles 1 and 11 (the reference piles) during 
the first load test, conducted 15 days after the completion of driving. 
It should be recalled that Pile 11 was subjected to a quick test, whereas 
Pile 1 was subjected to a standard one-hour-increment test. It is 
observed that both piles failed by plunging, followed by relaxation as 
further deformation developed. Failure in each case occurred abruptly, 
after a near-linear response, at a butt settlement of about 0.15 in. (3.8 
mm). This type of "brittle" failure was typical of all failures in 
compression for both group and reference piles throughout the test 
program . 

The tip loads depicted in Fig. 2.1 are based on pretest zeros and 
do not reflect the residual loads present prior to loading. Complete 
failure at the tip occurred at a downward movement of the tip of about 
0.2 in. (5 mm), or about 2 per cent of the pile diameter. (All "tip" 
loads referred to in this report are actually loads measured by a strain 
circuit approximately 1 ft. (0.305 m) above the tip.) It should be 
noted that the tip deformation required to mobilize full end bearing 
capacity exceeded the relative deformation between the shaft and the 
soil needed to mobilize maximum shaft resistance. Because of pile 
flexibility, shaft failure preceded tip failure. Since some relaxation of 
shaft resistance occurred after shaft-soil failure, the peak pile capacity 
was less than the sum of the peak shaft and tip capacities. 



45 




o 

z 

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46 



Pile 11 failed at an applied load of about 35 kips (156 kN) less 
than Pile 1. This may be due to the effects of undissipated pore water 
pressures that may have been generated when Pile 11 was driven in a 
partially water-filled pilot hole. Because of the early failure of Pile 11, 
the only comment that can be made on the comparison of the results of 
quick and standard tests is that both yielded nearly identical load- 
settlement curves in the working load range. The data suggest that 
factors such as the condition of the pilot hole are probably more 
important than the effects of testing method (for the two methods 
considered) in the soil at the test site. It is noted, however, that no 
basis exists for excluding Pile 11 data from the reference baseline, 
since some group piles may have behaved more like Pile 11 than Pile 1. 

The butt and tip load-settlement curves for the second load tests 
on the reference piles, conducted 78 days after completion of driving, 
are shown in Fig. 2.2. That figure also shows the butt load-settlement 
curves for Test 1 to provide a basis for comparison. Again, the tip 
curves are based upon pretest zeros. It is evident that both the butt 
and tip load-settlement behavior were more nearly identical between the 
two reference piles by this time and that appreciable apparent set-up 
had occurred. Plunging occurred at a butt settlement of about 0.2 in. 
(5 mm), and maximum tip load was also developed at a tip displacement 
of about 0.2 in. (5 mm). 

Figure 2.3 depicts the butt and tip load-settlement curves for the 
final reference pile load tests, conducted 105 days after completion of 
driving. Tip curves are again based on pretest zeros. That figure 
also shows the peak butt loads developed in the previous two tests. 
Note that at this time the capacity of Piles 1 and 11 were almost 
identical and that some loss in capacity was actually observed between 
Tests 2 and 3 for Pile 1. 

Cumulative load-settlement curves for the reference piles are given 
in Fig. 2.4. 

Groups . Load- settlement curves for the first 9-pile group test are 
given in Fig. 2.5. The load axis does not include the weight of the 
pile cap, which was 57 kips (254 kN). This figure compares the 
results of readings taken 5, 30, and 55 minutes after each loading 
increment was applied. The differences were insignificant up to about 
75 per cent of the plunging failure load. The pile cap experienced 
considerable tipping toward the north (Piles 8,9, and 10) as failure 
approached. This phenomenon is illustrated in Figs. 2.6-2.9, which 
diagram the attitude of the cap at various stages of loading. Separate 
load-settlement curves have therefore also been plotted on Fig. 2.5 for 
the north and south rows of piles. A maximum differential settlement of 
approximatley 0.3 in. (8 mm) was experienced across the piles at 
maximum applied load. Previous calculations had indicated that further 
rotation of the cap could induce plastic hinges in the piles at the base 
of the pile cap. The test was therefore terminated at this time. It is 
evident that in a gross sense (i.e., average pile settlement vs. 
average pile load, as depicted on Fig. 2.5) failure of the plunging type 

47 ~ 






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DISPLACED 
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\ WW WW \ 


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1 




SOUTH ELEV . 

DISPLACEMENT 
SCALE 

H H 

" 0.5" ' 
NOMINAL LOAD = 300 TONS 



FIGURE 2.6. CAP MOVEMENTS AT APPROXIMATELY ONE-HALF OF FAILURE LOAD, 
TEST 1 (1 ton = 8.9 kN; 1 in = 25.4 nun) 



52 



t 



N 



DISPLACED 
POSITION " 



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POSITION / 


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■% 





SOUTH ELEV. 

DISPLACEMENT 
SCALE 

h H 

0.5" ' 

NOMINAL LOAD = 600 TONS 



FIGURE 2.7. CAP MOVEMENTS AT APPROXIMATELY 90 PER CENT OF FAILURE LOAD, 
TEST 1 (1 ton =8.9 kN; 1 in = 25.4 mm) 



53 



DISPLACED 
POSITION' 




SNWNNWS 



PLAN VIEW 



^ 



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SOUTH ELEV. 



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UJ 



DISPLACEMENT 
SCALE 

H H 

r 0.5" ^ 

NOMINAL LOAD = 700 TONS 
(FAILURE) 



FIGURE 2.8. CAP MOVEMENTS AT FAILURE, TEST 1 (1 ton = 8.9 kN; 

1 in = 25.4 mm) 



54 




PLAN VIEW 




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sk > x \ WW 



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DISPLACEMENT 
SCALE 

FULLY UNLOADED 



FIGURE 2.9. CAP MOVEMENTS UPON REMOVAL OF LOAD, TEST 1 
(1 ton = 8.9 kN; 1 in = 25.4 nun) 



55 



occurred in the group at the maximum applied load. However, 
examination of the individual pile load-settlement curves (Figs. 
2.10-2.12) reveals that, because of the cap rotation, all of the piles did 
not fail simultaneously. Clearly, Piles 2,3,7,8,9, and 10 (on the central 
and northern rows) had failed by the time maximum load was reached, 
and Pile 6, on the southern row was in a state of impending failure. 
Piles 4 and 5 , also on the southern row, had not failed but were 
apparently near failure, based on observations of load-settlement curves 
of piles that had failed. The gross failure pictured in Fig. 2.5, then, 
is the result of the achievement of plunging failure in some piles followed 
by relaxation, while other piles (notably 4 and 5) were still attracting 
load. Therefore, even though the group failed in a gross sense (could 
not carry more load without settlement of the center of the cap 
increasing in an unlimited way), complete failure of every pile was not 
achieved in the first test. 

Complete tip failure was achieved in only a few of the piles. The 
locked-in tip loads at the end of the first group test were lower than 
those for the reference piles, presumably due to incomplete tip failure. 

The characteristic shape of the 5-minute load- settlement curve for 
the group, as well as the shapes of the load-settlement curves for the 
individual group piles, is similar to that for the reference piles, except 
for the greater deflections required to mobilize failure in the group and 
for the lack of relaxation or "deformation softening" in the group load- 
settlement curve. The latter effect was produced by non-simultaneous 
failures of individual piles that were themselves deformation softening. 

Considerably more attention will be directed toward the comparison 
of reference pile and group behavior later in this chapter. 

Load-settlement curves for the second and third 9-pile group tests 
are shown in Fig. 2.13. Individual pile load- settlement curves and cap 
movement diagrams for these tests may be found in Appendices D and 
F, respectively, of this report. The failure load and load-settlement 
relationships for Tests 2 and 3 were nearly identical, although the 
group capacity in each test exceeded that in the first test by about 200 
kips (890 kN). 

Movement of the jacks approximately 3 in. (75 mm) to the south 
produced a more nearly vertical push in these tests. This resulted in 
complete failure for all piles, including complete tip failure. Because of 
the vertical push, all piles failed approximately simultaneously in these 
two tests, with the result that the deformation softening experienced by 
the individual piles can be observed in the group load-settlement curves. 

The entire load-settlement history of the three 9-pile group tests 
is shown in the cumulative load-settlement graph in Fig. 2.14. 



56 



-20 



20 40 



LOAD (K) 
60 80 100 



120 140 160 



0.1- 


*g-^.i 1 1 


— i 1 


1 1 1 

r l FT. ABOVE 
/ GROUND LEVEL 


0.2- 








z 


L__-l FT ABOVE 
\ PILE TIP 






1 0.3- 

5 

UJ 

_i 
It 
u 0.4 1 

CO •. 


1 






05- 






/ 


0.6- 






----_' 




160 



---/ 



-20 



20 40 



LOAD (K) 
60 80 



100 120 140 160 



1 i 


"se^z" -1 — 


i i i i i 

- — ._- A FT ABOVE GROUND LEVEL 


- - 1 — 1 


0.1 




•"* — 
,. — -1 FT ABOVE TIP ^>x. 




■ SETTLEMENT (IN.) 
o o o 








0.5 


- ^^~~-- 






06 









FIGURE 2.10. LOAD- SETTLEMENT CURVES FOR INDIVIDUAL PILES ON NORTH ROW, 
TEST 1 (TOP: PILE 10; MIDDLE: 9; BOTTOM: 8) 
(1 k = 4.45 kN; 1 ft = 0.505 m; 1 in = 25.4 mm) 



57 



•20 




0-6 L 



LOAD (K) 
60 80 100 120 140 




160 



NOTE: UNCORRECTED FOR 

MOVEMENT OF REFERENCE 

SYSTEM 

TIP MOVEMENTS ARE FROM 

INTEGRATED STRAIN GAGE READINGS 



0.6 



120 140 160 




0.6 L 



FIGURE 2.11. LOAD- SETTLEMENT CURVES FOR INDIVIDUAL PILES ON CENTER ROW, 
TEST 1 (TOP: PILE 3; MIDDLE: 2; BOTTOM: 7) 
(1 k = 4.45 kN; 1 ft = 0.305 m; 1 in = 25.4 mm) 

58 



LOAD(K) 
60 80 100 




LOAD (K) 
20 40 60 80 



100 120 140 




0.6 <- 



LOAD (K) 
60 80 100 




06 L 



FIGURE 2.12. LOAD-SETTLEMENT CURVES FOR INDIVIDUAL PILES ON SOUTH ROW, 
TEST 1 (TOP: PILE 4; MIDDLE: 5; BOTTOM: 6) 
(1 k = 4.45 kN; 1 ft = 0.305 m; 1 in = 25.4 mm) 

59 



GROUP LOAD (K) 
600 800 1000 



* 



1200 1400 1600 



UJ 



UJ 
V) 




PLUNGING 
^LOAD FOR 
TEST 2 *** 



TEST 2 
22 JAN 80 



TEST 3 
9 FEB 80 



* AS INDICATED BY LEVEL I STRAIN GAGE CIRCUITS - 
DOES NOT INCLUDE CAP WGT. 

** AVG. ON PLANE I FT ABOVE GROUND LEVEL (FOR 
DIAL GAGES) 

*** 1475 K REACHED WHEN PUMPING TO NEXT LOAD 
(INDICATED LOAD FROM LOAD CELLS) 

° III!. l\ AVG. OF ALL PILE DIAL GAGE READINGS, 

A ,tbl d 5 MIN. READINGS 

• TEST 3 J 

+ AVG. OF DIAL GAGES ON NORTH ROW (PILES 8,9,10) 

TEST 3, 5-MIN RDGS. 

x AVG. OF DIAL GAGES ON SOUTH ROW (PILES 4,5,6)- 

TEST 3, 5 -MIN RDGS. 



FIGURE 2.13. LOAD- SETTLEMENT CURVES FOR SECOND AND THIRD 9-PILE GROUP 
TESTS (1 k = 4.45 kN; 1 ft = 0.305 m; 1 in = 25.4 mm) 



60 



LOAD (K) 
200 400 600 800 1000 1200 1400 1600 




UJ 

2 

UJ 

-J 

H 
UJ 
c/) 

UJ 
> 

5 



3 



1.0 



2.0 - 



TEST 2 




FIGURE 2.14. CUMULATIVE LOAD-SETTLEMENT CURVE FOR 9-PILE GROUP TESTS 
(1 k = 4.45 kN; 1 ft = 0.305 m; 1 in = 25.4 mm) 



61 



The load-settlement curves for the subgroup tests are shown in 
comparison with the third 9-pile test curve in Fig. 2.15. Relative 
capacity in each test was approximately in direct proportion to the 
number of loaded piles, and relative settlement at any given sub-failure 
load was approximately inversely proportional to the number of loaded 
piles. Detailed load- settlement plots for the individual piles in the 
subgroup tests can be found in Appendix D, and diagrams of cap 
motion are given in Appendix F. 

Comparative behavior of the reference piles, subgroups and the 
9-pile group is depicted in a normalized fashion in Fig. 2.16. It can be 
seen in that figure that the settlement corresponding to a given average 
load per pile increases with an increasing number of loaded piles in the 
group. The differences in capacities between the three group tests 
shown in Fig. 2.16 probably reflect statistical effects of removal of 
certain "strong" piles from the group tested earlier, but they also 
indicate that some degradation of soil resistance occurred with each 
succesive test and that the interval of time between tests was too short 
to permit thixotropic or chemical healing of the soil fabric. 

Uplift Tests . The butt load-uplift curves for the two reference 
piles and four group piles subjected to individual uplift tests are shown 
in Fig. 2.17. Superimposed on that figure is the average load- 
settlement graph for the reference piles in their final compression test. 
Significant differences between the uplift and compression tests are 
evident: 

(1) The average failure load in uplift was approximately 112 kips 
(498 kN), compared with an average failure load in compression in the 
reference piles of approximately 178 kips (792 kN). Since the 
theoretical suction on the tips of the test piles is less than 1 kip (4.45 
kN), it is apparent that the differences in capacities between uplift and 
compression testing reflect the maximum true compression loads 
developed on the tips of the test piles when they were loaded in 
compression. In this regard the load-uplift curves for the pile tips, 
based on pretest zeros, are plotted in Fig. 2.18. Negative tip loads at 
failure in the order of 40 to 50 kips (178 to 223 kN) are observed. 
These are not true loads but represent the removal of residual loads 
that were present at the end of the last compression loading. Fig. 2.18 
also shows the average compressive tip load-settlement curve for the 
last loading of the reference piles, zeroed before the test. The sum of 
the maximum value on that curve and the average of the maximum 
(negative) values for the uplift curves for Piles 1 and 11 should yield 
the true compression tip capacity for the reference piles. 

(2) The load-uplift curves were much more nonlinear than the 
load- settlement curves. This, again, is an expression of the release of 
residual loads in the piles and possibly the alteration of soil fabric 
around the piles as the direction of shear stress was reversed . 



62 



GROUP LOAD(K)* 
600 800 1000 




9- PILE 
GROUP, 
TEST 3 
19 FEB 80 



* AS INDICATED BY LEVEL I STRAIN GAGE CIRCUITS-DOES NOT INCLUDE CAP WGT 
** AVG. ON PLANE I FT. ABOVE GROUND LEVEL (FOR DIAL GAGES) 
KKX30-MIN READING. ALL OTHERS ARE 5-MIN READINGS 

SOME MINOR TIPPING TO SOUTH AND EAST OCCURRED DURING 
THE 5- PILE SUBGROUP TEST AND THE 4- PILE SUBGROUP TEST 



FIGURE 2.15. LOAD-SETTLEMENT RELATIONSHIPS FOR SUBGROUP TESTS 
(1 k = 4.45 kN; 1 ft = 0.305 mj 1 in = 25.4 nun) 



63 



LOAD PER PILE (K) 
60 80 100 120 




NORMALIZED BUTT LOAD-SETTLEMENT CURVES 
FOR THIRD REFERENCE AND 9~PILE GROUP TEST 
AND FOR 5- AND 4 -PILE SUBGROUP TESTS 



FIGURE 2.16. NORMALIZED LOAD- SETTLEMENT RELATIONSHIPS 
(1 k = 4.45 kN; 1 in = 25.4 mm) 



64 



"^ w. 0> 0> C7> O* 
UJ UJ UJ UJ UJ LJ 

=! —i _j —J _j _j 

Q-CL Q_Q_ Q.Q. 



1 



AW 



I I g 



• /N v_* ««^ «_* /l a 




< 
Q 



3 



u-> 



2 

AS 

II 



CO 
(X 
I— I 

CO 

O 
i—i 

< 
eg 

s 

H 

Uh 

h- i 
-J 
a, 

i 

1 

E- 
E- 



UJ 
OS 

u 
i— i 



CNI) lJldfl 



65 



.0 



0.8 



AVG. POSITIVE 
TIP LOAD vs. 
SETTLEMENT, 
REF. PILES, 
TEST 3 



(a) 



PILE I (r) 

PILE II (r) 

-PILE 2 (g) 

PILE 9 (g) 

PILE 5 (g) 

PILE 4 (g) 




-40 -60 

TIP LOAD (K) 

(a) INDICATES 30 "MINUTE READING 

(g) INDICATES GROUP PILE 

(r) INDICATES REFERENCE PILE 



-100 



FIGURE 2.18. TIP LOAD-UPLIFT RELATIONSHIPS (Ik = 4.45 kN; 1 in = 25.4 mm) 



66 



(3) There was a higher degree of variability of load-uplift 
response than of load-settlement response. 

Distribution of Loads to Piles 

The distribution of applied loads to the various piles in the first 
9-pile group test and pile settlements are depicted graphically in Figs. 
2.19 and 2.20 for several values of load. Graphical depictions for other 
tests may be found in Appendix F. Tables 2.1 and 2.2 summarize 
numerically the pile head load distribution results for all group 
compression tests. It is apparent that the loads remained relatively 
uniform throughout the working load range in both the 9-pile group and 
the subgroups. The corner piles received about 10 per cent more load 
than the center pile, and the edge piles received an intermediate value 
in the first (virgin) 9-pile test. These loads do not include the effect 
of the cap weight. 

At failure in the first 9-pile test the center pile attracted the 
largest load in the group. This effect could be partly due to the 
progressive nature of pile failure for this test, such that at the precise 
time the pile gages were read peak resistance may have existed in Pile 
2 but not in the other piles, and not a general phenomenon to be 
associated with the failure of pile groups. Subsequent tests did not 
indicate that the center pile attracted higher loads at failure than the 
remaining piles. 

Because of the tilting mechanism described earlier and because 
the piles did not fail simultaneouly , considerable variations in the pile 
loads existed when the group was unloaded after Test 1. Tensile loads 
existed on the exterior piles that failed first (8,10,7,3), and significant 
compressive load remained on the center pile (2) and on Piles 4 and 5, 
which did not fail. Piles 9 and 6 had relatively small compressive loads 
upon removal of the applied load. 

Similar variations can be observed for the other group and sub- 
group tests in Appendix F. 

Variation of Capacity with Time 

The variation of the average peak capacity of the reference piles 
and the average gross peak capacity of the 9-pile group for the three 
sets of compression tests are shown as functions of time plotted on a 
logarithmic scale in Fig. 2.21. There exists an apparent set-up in both 
the reference piles and the group, with set-up occurring at 
approximately the same rate in each system. The total set-up between 
the first and third tests, in the order of 30 kips (130 kN) per pile, is 
approximatley equal to the difference in average residual tip load 
between the post-drive condition (Figs. 1.22 and 1.23) and the last 
compressive loading, suggested by the negative tip load values in Fig. 
2.18. This fact implies that the capacity increase was not a result of 



67 




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69 



TABLE 2.1. DISTRIBUTION OF LOADS TO PILE HEADS: 
9-PILE GROUP TESTS (1 k = 4.45 kN) 



NOMINAL 


PILE 


PILE HEAD LOAD 


(k) 


A 


l/CDAPP E 


tit uc An rAAn f\*\ 








VtKAoC riLC ncrtu Livnu {*.j 


GROUP 
LOAD (k) 




TEST 1 


TEST 2 


TEST 3 




1 


2 


20.8 


16.6 


18.8 


CE 1: 


CENTER (TEST 1): 20.8 (10.3) 




3 


23.7 


18.1 


19.7 


ED 1: 


EDGE (TEST 1): 22.4 (11.1) 




200 


4 


22.7 


22.6 


22.9 


CO 1: 


CORNER (TEST 1): 23.0 (11.4) 




5 


21.6 


20.8 


21.9 


CE 2: 


CENTER (TEST 2): 16.6 ( 9.5) 




6 


21.7 


23.5 


24.6 


ED 2: 


EDGE (TEST 2): 18.9 (10.8) 




7 


21.6 


19.1 


19.8 


CO 2: 


CORNER (TEST 2): 20.8 (11.9) 




8 


23.9 


18.1 


19.7 


CE 3: 


CENTER (TEST 3): 18.8 (10.3) 




9 


22.5 


17.4 


17.6 


ED 3: 


EDGE (TEST 3): 19.8 (10.9) 




10 


23.5 


18.9 


16.7 


CO 3: 


CORNER (TEST 3): 21.0 (11.6) 




Sum 


201.8 


174.8 


181.8 








2 


40.2 


34.6 


37.6 




CE 1 


40.2 (10.4) 




3 


45.6 


39.9 


42.0 




ED 1 


42.8 (11.0) 


400 


4 


44.9 


46.8 


43.6 




CO 1 


44.3 (11.4) 




5 


41.3 


42.9 


41.8 




CE 2 


34.6 ( 9.6) 




6 


42.4 


45.5 


47.1 




ED 2 


39.4 (10.9) 




7 


40.6 


37.9 


39.6 




CO 2 


42.5 (11.7) 




8 


44.2 


38.0 


42.0 




CE 3 


37.6 (10.2) 




9 


43.6 


36.7 


37.9 




ED 3 


40.3 (11.0) 




10 


45.7 


39.7 


36.2 




CO 3 


42.2 (11.5) 




Sum 


388.3 


362.1 


367.6 








2 


59.9 


56.3 


59.8 




CE 1: 59.9 (10.3) 




3 


68.6 


67.2 


67.2 




ED 1: 64.0 (11.0) 


600 


4 


65.7 


74.4 


'67.9 




CO 1: 66.4 (11.4) 




5 


62.2 


70.4 


67.3 




CE 2: 56.3 ( 9.6) 




6 


64.3 


73.9 


74.1 




ED 2: 64.5 (11.0) 




7 


61.0 


61.8 


62.3 




CO 2: 67.4 (11.5) 




8 


67.5 


60.9 


65.0 




CE 3: 59.8 (10.2) 




9 


64.1 


58.6 


59.4 




ED 3: 64.1 (11.0) 




10 


68.0 


60.5 


61.0 




CO 3: 67.0 (11.5) 




Sum 


581.3 


583.9 


583.9 








2 


81.5 


75.0 


81.8 




CE 1: 81.5 (10.4) 




3 


92.9 


86.7 


89.1 




ED 1: 86.3 (11.0) 


800 


4 


89.6 


98.5 


91.8 




CO 1: 90.0 (11.4) 




5 


84.7 


92.9 


90.5 




CE 2: 75.0 ( 9.7) 




6 


86.5 


98.0 


98.0 




ED 2: 84.6 (11.0) 




7 


81.0 


81.6 


84.3 




CO 2: 89.6 (11.6) 




8 


90.1 


80.4 


87.0 




CE 3: 81.8 (10.4) 




9 


86.6 


77.1 


80.1 




ED 3: 86.0 (11.0) 




10 


93.7 


81.3 


82.6 




CO 3: 89.9 (11.5) 




Sum 


786.6 


771.4 


785.0 







1 ( ) indicates per cent of total , 



70 



TABLE 2.1. DISTRIBUTION OF LOADS TO PILE HEADS: 
9-PILE GROUP TEST (CONT'D) (1 k = 4.45 kN; 1 in. = 25.4 nun) 







PIJ.F HEAD LOAD 


(k) 






NOMINAL 


PILE - 






AVERAGE 


PILE HEAD LOAD (k) 


GROUP 
LOAD (k) 




TEST 1 


TEST 2 


TEST 3 






2 

7 


102.3 
106.0 


94. y 
108.6 


1UU.2 
112.4 


Lb 1 
ED 1 


1U2.3 (10.5) 
104.9 (10.8) 




J 


1000 


4 


112.0 


124.3 


115.3 


CO 1 


112.1 (11.6) 




5 


105.2 


116.7 


111.7 


CE 2 


94.9 ( 9.7) 




6 


106.8 


122.6 


122.8 


ED 2 


106.6 (10.9) 




7 


99.3 


101.8 


105.0 


CO 2 


113.3 (11.6) 




8 


113.7 


101.4 


109.2 


CE 3 


100.2 (10.2) 




9 


109.2 


99.2 


99.2 


ED 3 


107.1 (10.9) 




10 


115.9 


105.0 


103.4 


CO 3 


112.7 (11.5) 




Sum 


970.3 


974.3 


979.1 








2 


128.8 


117.0 


121.9 


CE 1 


128.8 (11.0) 




3 


124.6 


132.9 


136.7 


ED 1 


126.1 (10.8) 


1200 


4 


132.8 


151.3 


140.3 


CO 1 


133.2 (11.4) 




5 


127.3 


144.2 


139.2 


CE 2 


117.0 ( 9.8) 




6 


130.5 


148.4 


148.2 


ED 2 


130.3 (11.0) 




7 


116.6 


123.7 


127.9 


CO 2 


137.5 (11.6) 




8 


131.8 


124.2 


132.4 


CE 3 


121.9 (10.2) 




9 


135.8 


120.5 


121.7 


ED 3 


131.4 (11.0) 




10 


137.8 


126.2 


128.3 


CO 3 


137.3 (11.5) 




Sum 


1166.1 


1188.3 


1196.6 








2 


154.4 


135.7 


141.2 


CE 1 


154.4 (12.1) 




3 


135.1 


154.3 


159.6 


ED 1 


138.2 (10.8) 


1400 


4 


140.9 


171.4 


163.3 


CO 1 


141.9 (11.1) 




5 


139.4 


170.4 


162.4 


CE 2 


135.7 ( 9.8) 


(Failure 


6 


144.9 


174.2 


172.4 


ED 2 


151.6 (11.0) 


for Test 1- 


7 


126.0 


141.4 


144.5 


CO 2 


160.4 (11.6) 


Settlement= 


8 


131.0 


144.8 


153.0 


CE 3 


141.2 (10.2) 


0.295 in.) 


9 


152.2 


140.4 


143.3 


ED 3 


152.5 (11.0) 




10 


150.9 


151.2 


150.7 


CO 3 


159.9 (11.5) 




Sum 


1274.7 


1383.8 


1391.0 








2 


153.2 






CE 1 


153.2 




3 


135.3 






ED 1 


138.9 


1400 


4 
5 


147.1 
144.7 






CO 1 


141.5 


(May 


6 


145.6 










Settlement 


7 


125.2 










for Test 1 


8 


131.2 










= 0.420 


9 


150.4 










in.) 


10 

Sum 


142.1 
1274.9 











71 



TABLE 2.1. DISTRIBUTION OF LOADS TO PILE HEADS: 
9-PILE GROUP TEST (CONT'D) (1 k = 4.45 kN; 1 in. = 25.4 mm) 







PILE HEAD LOAD 


(k) 






NOMINAL 


PILE 






AVERAGE 


PILE HEAD LOAD (k) 


GROUP 
LOAD (k) 




TEST 1 


TEST 2 


TEST 3 






2 
3 




144.0 


153.4 






1600 




155.6 


173.6 








4 




145.7 


176.8 






(Failure for 


5 




158.5 


176.5 


CE 2 


144.0 (10.1) 


Test 2-Sett= 


6 




174.8 


181.5 


ED 2 


158.5 (11.1) 


0.48 in.) 


7 




149.1 


155.3 


CO 2 


162.3 (11.4) 




8 




169.9 


159.4 


CE 3 


153.4 (10.3) 


(Failure for 


9 




170.9 


150.0 


ED 3 


163.9 (11.0) 


Test 3-Sett= 


10 




158.6 


157.5 


CO 3 


168.8 (11.4) 


0.33 in.) 


Sum 




1427.0 


1483.9 






1600 


2 
3 




131.4 
143.1 


137.6 
159.6 






(Max. Sett. 


4 




139.9 


155.0 






for 5 min 


5 




151.4 


152.2 


CE 2 


131.4 ( 9.9) 


RDG for Test 


6 




159.6 


158.3 


ED 2 


146.7 (11.1) 


2=1.29 in.) 


7 




141.8 


138.9 


CO 2 


150.9 (11.4) 




8 




153.6 


147.4 


CE 3 


137.6 (10.3) 


(Max. Sett. 


9 




150.6 


136.6 


ED 3 


146.8 (11.0) 


for 5 min 


10 




150.4 


145.8 


CO 3 


151.6 (11.4) 


RDG for Test 


Sum 




1321. ':■ 


1331.3 






3=1.29 in.) 
















2 


24.2 


- 2.3 


2.6 


CE 1 


24.2 





3 


- 9.2 


-12.0 


3.4 


ED 1 


-4.6 




4 


6.8 


-17.0 


-0.4 


CO 1 


-5.5 


Unloaded 


5 


14.7 


4.7 


2.4 


CE 2 


-2.3 




6 


0.6 


- 0.2 


-3.0 


ED 2 


0.5 




7 


-12.7 


0.6 


-2.5 


CO 2 


-2.3 




8 


-17.9 


4.5 


-1.4 


CE 3 


2.6 




9 


4.1 


8.5 


-1.2 


ED 3 


-1.6 




10 


-11.3 


3.5 


-1.6 


CO 3 


-0.7 




Sum 


- 0.8 


- 9.8 


-1.6 







NOTE: Failure in Test 1 produced tendency for equalization of loads. 
Failure in Test 2 did not. Possibly due to progressive fail- 
ure effects induced by cap rotation at failure, combined with 
lack of ability to read gages on all piles simultaneously. 



72 



TABLE 2.2. DISTRIBUTION OF LOADS TO PILE HEADS: 
SUBGROUP TESTS (1 k = 4.45 kN) 



NOMINAL 

GROUP LOAD (k) 


PILE 


PILE HEAD LOAD (k) 


AVERAGE PILE HEAD LOAD (k) 


5- PILE SUBGROUP 


4 -PILE SUBGROUP 


100 


2 


19.4 





CE 5: 19.4 (18. 8) 1 




3 


19.3 


25.7 


(Center, 5 Pile Group) 




5 


17.8 


28,8 


CO 5: 21.0 (20.3) 




7 


23.1 


25.0 


(Corner, 5 Pile Group) 




9 


23.7 


18.7 


CO 4: 24.6 (25) 




Sum 


103.3 


98.2 


(Corner, 4 Pile Group) 


200 


2 


30.5 





CE 5: 30.5 (17.8) 




3 


32.5 


55.0 


CO 5: 35.1 (20.5) 




5 


34.2 


55.4 


CO 4: 52.6 (25) 




7 


39.1 


51.9 






9 


34.7 


47.9 






Sum 


170.9 


210.2 




300 


2 


50.5 





CE 5: 50.5 (18.4) 




3 


58.2 


76.5 


CO 5: 55.9 (20.4) 




5 


55.9 


75.4 


CO 4: 71.9 (25) 




7 


56.2 


71.2 






9 


53.4 


64.7 






Sum 


274.2 


287.8 




400 


2 


69.3 





CE 5: 69.3 (18.6) 




3 


77.9 


106.8 


CO 5: 75.9 (20.4) 




5 


76.6 


108.9 


CO 4: 102.4 (25) 




7 


75.6 


99.7 






9 


73.5 


94.2 






Sum 


372.9 


409.6 




500 


2 


84.7 





CE 5: 84.7 (18.7) 




3 


93.1 


131.7 


. CO 5: 92.2 (20.3) 




5 


94.0 


134.0 


CO 4: 126.4 (25) 




7 


93.7 


123.5 






9 


88.1 


116.5 






Sum 


453.7 


505.7 




600 


2 


106.7 





CE 5: 106.7 (18.4) 




3 


119.6 


143.1 


CO 5: 118.5 (20.4) 




5 


122.1 


146.8 


CO 4: 137.7 (25) 




7 


119.1 


135.4 






9 


113.0 


125.4 






Sum 


580.4 


550.7 





1 ( ) indicates per cent of total. 



73 



TABLE 2.2. DISTRIBUTION OF LOADS TO PILE HEADS: SUBGROUP TESTS (CONT'D) 

(1 k = 4.45 kN) 



NOMINAL GROUP 
LOAD (k) 


PILE 


PILE HEAD 


LOAD (K) 


AVG. PILE HEAD LOAD (k) 


5-PILE SUBGROUP 


4-PILE SUBGROUP 


700 


2 


125.1 




CE5: 125.1 (18.3) 




3 


141.5 


151.9 


C05: 139.3 (20.4) 




5 


144.0 


153.3 


C04: 145.2 (25) 




7 


137.6 


141.6 






9 


134.1 


133.9 






SUM 


682.2 


580.7 




800 


2 


142.1 




CE5: 142.1 (18.3) 




3 


166.6 


157.0 


C05: 158.2 (20.4) 




5 


167.0 


158.9 


C04: 151.0 (25) 




7 


149.5 


148.7 






9 


149.6 


139.6 






SUM 


774.9 


604.1 







2 


-2.2 




CE5: -2.2 


(unloaded) 


3 


7.0 


2.0 


C05: 1.1 




S 


2.6 


-1.1 


C04: 0.0 




7 


-5.0 


-0.2 






9 


-0.4 


-0.6 






SUM 


2.0 


0.1 





74 



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75 



increased side resistance but rather a function of increasing tip 
capacity produced by cyclic loading of the soil beneath the pile tips. 
The residual tip loads which developed after each test apparantly 
consolidated and strengthened the soil beneath the pile tips, rendering 
an increased total tip capacity (based on the unloaded condition, not 
the pretest condition) for each successive loading. 

The center pile, Pile 2, maintained an approximately constant peak 
capacity in the presence of increasing residual tip load, indicating that 
average usable side resistance actually decreased with each successive 
loading. Further consideration to timewise decreases and increases in 
unit side resistance in the various soil layers is given in Chapter 3. 

It must be concluded that the results of multiple load tests on the 
same pile or pile group in this soil (a strain-softening, rapidly draining, 
over consolidated clay) do not yield accurate information on the gain of 
capacity with time that would be experienced by a typical pile or pile 
group, which would normally experience failure only once. 

Settlement Ratios 

One useful means of depicting group action is to express the ratio 
of butt settlement of a group of piles at a given average load per pile 
to that of a single, isolated pile under the same load. This ratio, 
termed the settlement ratio, describes how much more the group will 
settle than the isolated pile under similar loading conditions (short-term 
static loading in the case of this study). Settlement ratios at various 
percentages of group failure load (cap weight neglected) are shown 
graphically in Fig. 2.22 for the three 9-pile group tests. Settlements 
were considered at 1 ft. (0.305 m) above the ground surface for all 
piles, and the appropriate isolated pile settlement was taken as the 
average reference pile settlement for the particular set of tests (first, 
second, or third) being considered. The three sets of tests yielded 
consistent settlement ratios over a wide range of loads except for a low 
value at the low end of the. load scale for Test 1. This low value is 
believed to be due to experimental errors associated with the very small 
settlements encounted at low load values and not to any physical 
phenomenon. It is also observed that the settlement ratio was 
essentially constant in all tests over a wide range of loads. 

Errors associated with making settlement readings are discussed 
and evaluated in Appendix E. The ratios depicted in Fig. 2.22 and in 
the figures and tables which follow are based on as-read settlement 
readings. These readings are believed to be essentially correct for the 
reference piles but to be slightly low for the group piles due to small 
movements in the 40 ft. (12.2 m) long reference system produced by 
strains induced in the soil by the loaded piles. However, theoretical 
considerations, as well as back-up readings made using the microhead 
level, indicate that the reported settlement ratios are probably no more 
than 20 per cent too low. 



76 



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Figure 2.23 shows the average settlement ratios for the three 9-pile 
tests and the settlement ratios measured for the 5-pile and 4-pile 
subgroup tests. These latter ratios are based on the set of reference 
pile tests associated with the third 9-pile group test. Except at very 
low loads the settlement ratios for the smaller groups are lower than for 
the 9-pile group, in the order of 1.2 to 1.3, with the 4-pile group 
yielding the smallest values. It should be recalled that the nominal 
center-to-center spacing of the loaded piles was 3 diameters for the 
5-pile group and 4.2 diameters for the 4-pile group; hence, Fig. 2.23 
should be understood to represent the effects of both the number of 
loaded piles and of pile spacing. It is the authors' opinion that the 
loading history effects discussed in the previous section were relatively 
minor between the third 9-pile test and the final subgroup test, so that 
the settlement ratios in Fig. 2.23 represent a valid comparison. 

Figure 2.23 also shows the settlement ratios that are predicted for 
the various groups by the elastic solid model (see Interim Report ) , 
assuming flexible piles and assuming two depthwise variations of soil 
modulus for an incompressible soil. It is evident that the "Gibson" soil 
(Young's modulus increasing linearly from zero at the soil surface to 
the modulus indicated by the pressuremeter at the pile tips) yields 
results closer to the measured values than does representation of the 
soil as a semi-infinite halfspace (constant modulus equal to the 
pressuremeter value at the mid-depths of the piles), although both soil 
representations yield settlement ratios that are consistently too high. 
This observation suggests that the stiffness of the soil in the zone 
beneath the pile tips may be more important than that above the pile 
tips in controlling short-term settlements and that the reinforcing effect 
of the piles, not considered by the elastic solid or hybrid models, may 
effectively stiffen the load -settlement response. For these reasons it 
appears that selection of soil moduli that are in excess of the in-situ 
modulus determined through high quality field or laboratory tests in the 
soil zone from two to three group widths below the pile tips would be 
most appropriate as inputs to the hybrid model. 

The settlement ratios for the pile tips were greater than those for 
the pile butts, as indicated for Test 1 in Table 2.3. Common reference 
and group pile tip loads for making the calculations did not include 
residual loads, so that the ratios with respect to true, or absolute, 
loads would be slightly higher than those reported in Table 2.3. These 
data suggest that group action influenced tip settlements to a greater 
extent than it influenced pile compression. 

Induced Settlements 

Performance of the subgroup tests provided an opportunity to 
assess the effects of the loaded piles on the unloaded piles that had 
been detached from the pile cap. Settlement effects are described 
here, while load transfer effects are described in Chapter 3. The 
measured induced effects give some insight into mechanical group 

78 



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80 



behavior and provide data for futher fundamental investigations of 
mathematical models. 

Table 2.4 summarizes the butt settlements of the unloaded corner 
piles (outside of the subgroups of loaded piles: Piles 4,6,8, and 10 
from the 9-pile group) as ratios of butt settlements of the loaded piles 
(Piles 3,5,7, and 9) for various values of applied load in both the 4- 
and 5-pile subgroups. Likewise, the ratios of the settlement of the 
unloaded center pile (Pile 2) in the 4-pile test to the average settlement 
of the loaded piles is also shown. The unloaded corner piles can be 
seen to have settled somewhat more in the 5-pile test than in the 4-pile 
test for values of load less than about 75 per cent of failure of the 
loaded group. The unloaded center pile can also be seen to have 
settled more than the unloaded corner piles in the 4-pile test. These 
results are not unexpected, but the magnitudes of these effects are 
relatively smaller than can be inferred by most elastic solid model 
algorithms . 

If the mean settlements of the unloaded corner piles are plotted as 
functions of average load per pile applied to the loaded piles for both 
subgroup tests, the differences in the plots should approximately 
represent the settlement effect produced by the center pile (loaded in 
the 5-pile test but not loaded in the 4-pile test) on the typical corner 
pile 4.2 diameters away. The result would be essentially a two-pile 
elastic interaction effect except for the presence of the neighboring 
piles and the effects of load history. Such a plot is given in Fig. 2.24 
along with a graph of settlement differences versus single pile load. 
The latter relationship is essentailly linear up to about one-half of the 
failure load, whereupon induced settlement does not increase further 
with additional applied load. The lack of induced settlement response 
at higher loads may be a result of experimental error in measurning 
settlements, and a possible correction to the curve is suggested in Fig. 
2.24. 

Efficiencies 

A second principal means of expressing group action is by the 
ratio of the failure load of the pile group to the average failure load of 
the reference piles times the number of loaded piles in the group. This 
ratio is defined as the "efficiency" of the group. Efficiency is a 
characteristic often used by designers to size pile groups based on 
estimiated or measured capacities of single piles. The term "failure" 
throughout this section will refer to the plunging load of a pile or pile 
group . 

Table 2.5 presents a comprehensive summary of total peak 
capacities of every pile in each of the first three sets of tests (all 
reference pile tests and 9-pile group tests). It also gives peak side, 
or shaft, capacities and maximum tip loads measured on every pile. 
Based on the average peak shaft, tip, and total capacities for the two 



81 



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5" PILE 
SUBGROUP 





IMPLIED 

SETTLEMENT INDUCED IN CORNER 
PILES (4,6,8,10) DUE TO LOADING 
A CENTER PILE (2) (4.2 DIAMS. 
AWAY) 




4-PILE SUBGROUP 



60 80 

LOAD /PILE (K) 



POSSIBLE 
CORRECTION 



100 



120 




40 60 

SINGLE PILE 



80 
LOAD(K) 



120 



FIGURE 2.24. SETTLEMENTS IN UNLOADED PILES VERSUS LOAD PER LOADED PILE (ABOVE); 
SETTLEMENT DIFFERENCES IN CORNER PILES BETWEEN 5- AND 4-PILE SUBGROUP TESTS 
(BELOW) (1 k = 4.45 kN; 1 in = 25.4 mm) 



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84 



reference pile tests that were associated with each group test, shaft, 
tip, and total efficiency for each group pile in each of the three 9-pile 
tests was calculated and tabulated in Table 2.5. The tabulated shaft 
and tip loads do not include the residual loads nor do they include the 
effect of cap weight. It is readily evident that the average shaft 
efficiency was slightly less than unity and the average tip efficiency 
was greater than unity, especially during the last tests. The high 
indicated tip efficiencies are believed to be partly due to lower residual 
tip loads in the group piles. The low indicated tip efficiency for Pile 7 
may be due to a measurement error associated with the presence of a 
quater bridge strain gage circuit at the tip of that pile. 

It should again be noted that the shaft and tip efficiencies relate 
to maximum loads that were not developed at compatible pile deflections, 
so that the total efficiency of any pile is less than the sum of the 
products of shaft efficiency and reference shaft capacity and tip 
efficiency and reference tip capacity. Furthermore, the peak total pile 
loads do not necessarily sum to the group capacity for a given test 
since the peak loads were not all developed simultaneously. 

Corresponding efficiency tabulations for the subgroup tests are 
presented in Table 2.6. Baseline reference pile values are those from 
the set of tests conducted prior to the third 9-pile group test. Shaft 
efficiencies were somewhat lower in the subgroup tests than in the 
9-pile group tests, possibly due to a lack of healing time between the 
final 9-pile test, the 5-pile test, and the 4-pile test. Tip efficiencies 
were generally lower for the 5-pile test than for the final 9-pile test 
and lower in the 4-pile test than in the 5-pile test. This effect is 
believed to be due to the continued buildup of residual tip loads on the 
piles subjected to the subgroup tests to the point where, in the final 
(4-pile) test, the pretest residual loads were at such a magnitude that 
the tip loads mobilized during load testing in excess of the pretest 
residual loads were generally less than the corresponding mobilized 
loads in excess of pretest residual loads for the reference pile tests, 
conducted prior to the third 9-pile test. 

Total or overall efficiency based on geometric position in the group 
is tablutated in Table 2.7. In general, the corner piles tended to be 
slightly more efficient than the edge piles, which, in turn, were slightly 
more efficient than the center pile. In the first 9-pile test, the center 
pile was the most efficient by a small margin. Table 2.7 also gives the 
average shaft and tip efficiencies for all group piles by test. The 
lowered efficiencies in the subgroup tests are due to statistical effects 
of removing slightly "stronger" piles from the original group and to the 
load history effects, including insufficient healing time, discussed 
earlier . 

Finally, the overall efficiencies for the five group tests are 
evaluated in Table 2.8. In this table efficiency has been defined in 
four ways which comprise all combinations of the following two criteria: 



85 



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86 



TABLE 2.7. OVERALL EFFICIENCY BY GEOMETRIC POSITION AND AVERAGE SHAFT 

AND TIP EFFICIENCIES 



OVERALL 
EFFICIENCY BASED ON 
GEOMETRIC POSITION IN GROUP 



(a) 



TEST 


CENTER 
PILE (2) 


AVG. OF 

EDGE 
PILES (3,5,7,9) 


AVG. OF 
CORNER 
PILES(4,6,8,I0) 


1 


1.04 


0.93 


096 


2 


0.84 


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0.98 


3 


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0.95 


5-PILE 
SUB- 
GROUP 


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089 


- 


4-PILE 
SUB- 
GROUP 


- 


085 


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AVERAGE 
SHAFT AND TIP 
EFFICIENCIES' ' 



TEST 


AVG. SHAFT 
EFFICIENCY 


AVG. TIP 
EFFICIENCY 


1 

2 

3 

5-PILE 
SUBGROUP 

4-PILE 
SUBGROUP 


0.93 
095 
0.86 

0.82 
0.85 


NOT ADEQUATELY 
DEFINED 

1.32 
1.40 

1.15 
0.81 



(a) NEGLECTING WGT. OF CAP AND LOADING ACCESSORIES 



87 



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88 



(1) failure defined by (a) load required to produce plunging failure of 
the center of the pile cap ("gross" failure), or (b) the sum of the 
plunging or maximum loads in each of the loaded piles comprising the 
group ("average" failure) and (2) failure load consisting of (a) applied 
load only, or (b) applied load plus the weight of the pile cap and 
loading accessories. For the condition of either average or gross 
failure, where the failure load included the weight of the cap, the 
overall efficiency was between 0.95 and 0.99 for the three 9-pile tests 
and was 0.92 to 0.93 for the subgroup tests. 

It can therefore be concluded that the efficiency of a full-scale 
pile group of 4 to 9 piles at this site is essentially unity. It should be 
emphasized that these results are in conflict with the various design- 
type efficiency models described in the Interim Report , which either 
assume that the group will fail in a "block" mode, thus enhancing 
efficiency (at this site; in other soil types efficiency can be less than 
unity with the block model), or that some capacity reduction factor 
should be applied to each pile based on geometric effects (e.g., Feld's 
Rule). This is not to argue that such effects would not exist in other 
soil profiles or in groups in the same soil with a significantly larger 
number of piles or at closer spacings than were employed in this study. 

Pore Water, Total and Effective Pressures Developed During Load Tests 

Synoptic tabulations of pore and total pressure readings made on 
the piles during the load tests are presented in Appendix F. Because 
of the sensitivity of the total pressure cells to temperature, documented 
in Appendix E, a means of correcting the total pressure readings also 
had to be developed. This procedure, which was partially subjective, 
is also described in Appendix F. All of the total pressure readings 
reported in this chapter are corrected readings. Since effective 
stresses were determined as the differences in total and pore water 
stresses, effective stress values are also "corrected" values. Pore 
water pressure values are the raw values that were read during the 
tests . 

Condiderable scatter occurred among the total pressure cells. The 
reader can readily observe the scatter in the stress tables in Appendix 
F, and some thoughts concerning the reasons for the scatter are offered 
in Appendix E. Unknown temperatures in the sensing fluid, surface 
irregularities on the piles in the vicinity of the pressure cells, the use 
of flat sensor faces (as opposed to the cylindrical surface of the piles), 
and soil property variations are thought to be the primary causes of 
the scatter. 

For this reason, it was not possible to discern differences in total 
or effective stress patterns on the various group piles. Therefore, 
data are presented in this chapter in terms of average total and 
effective stresses on the four group piles that were instrumented for 
lateral pressure and separately for the single reference pile (Pile 1) 
that was so instrumented. 

89 



Reference is made to Figs. 1.15-1.17, which indicated that 
significant excess pore pressures were developed during pile installation 
but that those pressures dissipated very rapidly thereafter to a level 
near the hydrostatic level by the beginning of the first set of load 
tests. Figure 1.18 depicted the general changes in lateral effective 
stresses at the pile surface produced by pile installation (but not by 
loading) in terms of effective earth pressure coefficients. Corrected, 
rather than raw, values of total pressure were used to produce the 
relationship shown in that figure. Measured lateral effective earth 
pressure coefficients after installation at the faces of the piles can be 
seen to be about 1.2 times the in-situ earth pressure coefficients at the 
9 and 19 ft. (2.7 and 5.8 m) levels and 2.4 times the in-situ values at 
the 34 and 41 ft. (10.4 and 12.5 m) levels. Very little 
difference between the average indicated effective stress on the group 
piles and measured effective stress on the reference piles existed at the 
upper level or at the lower two levels. No definitive statement can be 
made about the second level due to the unrealistically low pressure 
coefficient observed there on the reference pile. 

Attention is called to the fact that essentially no load transfer 
occurred over the top halves of the piles during the driving process 
(Figs. 1.9 and 1.10), yet approximately four days after driving (the 
time represented in Fig. 1.18) the effective lateral earth pressure 
coefficients in the top halves of the piles approximated the in-situ 
coefficients. From these measurements it is inferred that an annular 
space ("gap zone") may have developed between the pile and soil 
during driving over the top approximately one-half of the embedded 
portion of the pile at full penetration which later closed due to lateral 
expansion of the soil. 

Pressure Changes During Loading. Plots of total and pore 
pressure changes that occurred during load testing on representative 
lateral pressure cells at the first through fourth levels are shown in 
Figs. 2.25 through 2.28. Changes in pore pressures at the pile face 
are seen to have been small during loading, generally in the order of + 
2 psi (14 kN/m 2 ) or less. Some small negative changes were noted at 
the 41 ft. (12.5 m) depth. 

Total pressure changes were also generally small, although large 
changes can be observed after large relative movement occurred 
between the pile and soil for some cells, (e.g., Pile 4, 34 ft. (10.4 m) 
depth) These large changes, furthermore, were observed to be 
functions of the direction of loading, typically being positive for 
compression tests and negative for uplift tests. This phenomenon 
appears to have resulted from non-vertical alignment of the cells, 
produced by inadvertent battering of the piles and by initial placement 
of the cells on the piles such that the tops of the sensor plates pro- 
truded slightly beyond the bottoms. The need for this placement is 
discussed in Appendix E. Total stresses conesponding to failure were 
were taken as the values just preceding the large increases or 
decreases, where such large changes occurred. 

90 



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TOTAL AND PORE PRESSURE CHANGES 
DURING LOAD TESTING 



PILE NO. I 
9' DEPTH 




± 



x 



j_ 



50 100 150 200 

REFERENCE -PILE OR AVERAGE GROUP LOAD PER PILE (KIPS) 

50 100 150 200 

+ 8 



T 



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+ 4 



PILE NO. 2 
9' DEPTH 







O-- 

A— 
A-- 

CD- 
FIGURE 2.25. 



ARROW INDICATES TOTAL PRESSURE AT FAILURE AS 
USED FOR CORRELATIONS 

TOTAL PRESSURE CHANGE , TEST NUMBER I 

PORE PRESSURE CHANGE , TEST NUMBER I 

TOTAL PRESSURE CHANGE, TEST NUMBER 3 

PORE PRESSURE CHANGE, TEST NUMBER 3 

TOTAL PRESSURE CHANGE, UPLIFT TEST 

PORE PRESSURE CHANGE, UPLIFT TEST 

PORE AND TOTAL PRESSURE CHANGES ON PILES AT 9-FOOT (2.7 M) 
DEPTH (1 k = 4.45 kN; 1 psi = 6.89 kN/m 2 ) 



91 



TOTAL AND PORE PRESSURE CHANGES 
DURING LOAD TESTING 






o 



PILE NO. I 
19' DEPTH 




x 



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PILE N0.5 
9' DEPTH 




ARROW INDICATES TOTAL PRESSURE AT FAILURE AS USED FOR CORRELATIONS 

• TOTAL PRESSURE CHANGE, TEST NUMBER I 

O PORE PRESSURE CHANGE, TEST NUMBER I 

A TOTAL PRESSURE CHANGE, TEST NUMBER 3 

A PORE PRESSURE CHANGE, TEST NUMBER 3 

■ TOTAL PRESSURE CHANGE, UPLIFT TEST 



□ PORE PRESSURE CHANGE, UPLIFT TEST 



FIGURE 2.26. PORE AND TOTAL PRESSURE CHANGES ON PILES AT 19 -FOOT (5.8 M) 
DEPTH (1 k = 4.45 kN; 1 psi = 6.89 kN/m 2 ) 



92 



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PILE NO. 4 
34' DEPTH 




+12 








+ 8 






\J 




+ 4 






y/jl 





-4 


m __ . _ 




«s~*C 


i — & 


- 


7\ 






-8 










-12 


- 








-16 











ARROW INDICATES TOTAL PRESSURE AT FAILURE 
AS USED FOR CORRELATIONS 

TOTAL PRESSURE CHANGE, TEST NUMBER I 



O PORE PRESSURE CHANGE, TEST NUMBER I 

A TOTAL PRESSURE CHANGE, TEST NUMBER 3 

A PORE PRESSURE CHANGE, TEST NUMBER 3 

■ TOTAL PRESSURE CHANGE, UPLIFT TEST 

O pore PRESSURE CHANGE, UPLIFT TEST 

FIGURE 2.27. PORE AND TOTAL PRESSURE CHANGES ON PILES AT 34 -FOOT 
(10.4 M) DEPTH (1 k = 4.45 kN; 1 psi = 6.89 kN/m 2 ) 

93 



TOTAL AND PORE PRESSURE CHANGES 
DURING LOAD TESTING 









PILE NO. 1 






+ 8 
+4 












41' DEPTH 




LU 




^^-O 


O 





-*" ~~ <P ■ ■ p 


UJZ 




cr< 






ox 


-4 


A 


*£ 






O* 


-8 


— \ 


_|</> 






<w 


-\? 




|_IU 






otr 






l-CL 


-16 


\ 




-20 


i i 



50 



100 



200 



REFERENCE PILE OR AVERAGE GROUP LOAD PER PILE (KEDS) 
50 100 150 200 




ARROW INDICATES TOTAL PRESSURE AT FAILURE AS 
USED FOR CORRELATIONS 

• TOTAL PRESSURE CHANGE, TEST NUMBER I 

O PORE PRESSURE CHANGE, TEST NUMBER I 

▲ TOTAL PRESSURE CHANGE, TEST NUMBER 2 

A PORE PRESSURE CHANGE, TEST NUMBER 2 

B TOTAL PRESSURE CHANGE, UPLIFT TEST 



□ PORE PRESSURE CHANGE, UPLIFT TEST 

FIGURE 2.28. PORE AND TOTAL PRESSURE CHANGES ON PILES AT 41- FOOT 
(12.5 M) DEPTH (1 k = 4.45 kN; 1 psi = 6.89 kN/m 2 ) 

94 



It should be pointed out that the surface along which the pore 
pressures and total stresses were measured may not have been the 
actual failure surface, which may have developed at some slight distance 
into the soil. It is the opinion of the authors that since the pore 
pressure measurements were made very close to the failure surface in a 
saturated soil that the reported pore pressures developed during 
loading are very close to the values on the failure surface. The small 
changes in pore pressures along the shafts of the piles during loading 
suggest that the soil can possibly be treated as a frictional material for 
purposes of computing unit side resistance, and the small observed pore 
pressure changes coupled with the large observed increase in lateral 
effective earth pressure coefficient due to installation of the piles, in 
the bottom approximately one-fourth of the piles, suggests that the soil 
that was undergoing failure near the bottoms of the piles may have also 
been at or near the "critical state." This speculation is substantiated 
by the generally minor strain softening (relaxation) that occurred in 
the unit side shear-relative deformation (f-z) curves in the 30 to 40 ft. 
(9.2-12.2 m) depth range (described in Chapter 3 and in Appendix D). 
More significant strain softening occurred below 40 ft. (12.2 m), 
possibly due to a reduction in confining pressures caused by downward 
movement of the pile tips, and is not necessarily an indication that the 
soil undergoing failure was not at the critical state. The existence (or 
absence) of critical stress states is an important consideration for 
applying new advanced design procedures for pile capacities which are 
predicated on critical state soil mechanics theory. 

The absence of significant increases in lateral effective earth 
pressure coefficients observed at the top two levels of earth pressure 
cells (roughly top one-half of the piles, where the soil was highly 
overconsolidated) suggests that the failing soil may not have achieved 
the critical state at those levels. 

Horizontal Variations in Pore Water Pressure. Figures 2.29-2.34 
present graphical summaries of pore pressure variations (not excess 
pressures) along the profile lines described in Fig. 1.19. The purpose 
of these graphs is to provide a comparison between pore pressures 
measured on the surfaces of the piles and those measured in the soil 
mass and to show variations among the piles and through the soil. The 
single numeral following the letter P designates a pile piezometer (e.g., 
PI designates Pile 1, a reference pile, P2 designates Pile 2, a group 
pile, etc.), while a series of three numerals following the letter P 
designates a ground, or soil mass, piezometer (e.g., P345 designates a 
ground piezometer at a depth of 34 ft. (10.4 m) at position 5; see Figs. 
1.6 and 1.7). The pore pressures shown during testing are those 
values read five minutes after a load application. 

Figures 2.29-2.31 present pore water pressure readings prior to 
the beginning of each compression test. Figure 2.29 pertains to the 
reference pile tests and contains values only for the reference pile 
instrumented for lateral pressure (Pile 1); Figure 2.30 pertains to the 



95 



9- 



19- 



f34- 

UJ 
O 



41- 



50- 



10 
5 


15 

10 
5 





_ SECTION A-A 

Pi 



30 

_ 25 

C/> 20 

°- 15 

2 10 

or 

</> 
(f) 
UJ 

cr 

0. 30 

25 

20 

15 

10 

5 


40 
35 
30 
25 
20 
15 
10 
5 




SECTION D-D 



PI95 PI93 
"PI 



- i^\i 



LEGEND 

BEFORE TEST I (11/16/79) 

- — BEFORE TEST 2 ( 1/18/80) 

• — BEFORE TEST 3 (2/14/80) 



P345 



SECTION F-F 

P342 P34I 



P344 P343 




I 



(NOT 
(STABLE) 



SECTION 1-1 



Pi 



P503 



SECTION K-K 

P504 ' P502 P50I 




FIGURE 2.29. HORIZONTAL VARIATION IN PORE PRESSURE IN SOIL AND ON PILE 1 
PRIOR TO REFERENCE TESTS (1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ) 



96 



9" 



19" 



Ll. 

- 34 

i 
\- 

CL 
LLl 
Q 



41- 



50-J 



15- 
10 - 



SECTION B-B 




P4 



PI95 PI93 



in 

Q. 



cr 

CO 
CO 
UJ 
QC 



30 
25 
20 
15 
10 
5 




30 
25 
20 
15 
10 
5 


40 
35 
30 
25 
20 
15 
10 

5 




SECTION E-E 
P5 P2 



P3 



P4 




P345 



SECTION H-H 



P344 



P342 



P5 




P3 



P4 



P5 



SECTION J-J 
P2 



P3 



P4 




P504 



SECTION K-K 



P503 



P50I 




331 



LEGEND 



• BEFORE TEST 

NO.I (11/21/79) 

▲ BEFORE TEST 

N0.2( 1/22/8 0) 

■ BEFORE TEST 

N0.3 (2/18/80) 



FIGURE 2.30. HORIZONTAL VARIATION IN PORE PRESSURE IN SOIL AND ON GROUP 
PILES PRIOR TO 9-PILE TESTS (1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ) 



97 



9- 



9- 



l- 

X 

I- 

CL 

LlJ 
Q 



34- 



41 - 



50 J 



SECTION B-B P5 



4 - 
2 - 



30 
25 

0_ 15 

Z "0 

- 5 

uj o 

Z) 
CO 
CO 
UJ 

cr 



20 - 

15 - 

10 - 

5 - 



P2 




P3 



P4 



PI 95 



PI93 



SECTION E-E 




P4 



P345 SECTION H-H p 344 P3 42 (£343) 

" P4 



P5 




SECTION J-J 
P5 P2 



P3 



P4 




SECTION K-K 



P503 



P504 



P502 



P50I 




LEGEND 



■ BEFORE TEST 
NO. 3(2/18/80) 

BEFORE 5-PILE 
TEST(2/26/80) 

BEFORE 4-PILE 
TEST(2/29/80) 



FIGURE 2.31. HORIZONTAL VARIATION IN PORE PRESSURE IN SOIL AND ON GROUP 
PILES PRIOR TO SUBGROUP TESTS (1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ) 



98 



9 - 



19 - 



x 

Q. 



34- 



41- 



50 



10 r SECTIQN A-A 

PL 



7Z 35 
^30 

Z 25 
- 20 

UJ 15 

^ 10 

8 n 

UJ 

q: 
a. 



35 
30 
25 
20 
15 
10 
5 


40 

35 

30 

25 

20 

15 

10 

5 





SECTION D-D 

PI95 PI93 

PI 




LEGEND 

30 TON, TEST I 

FAILURE, TEST I 



40 TON, TEST 2 

FAILURE, TEST 2 



— 40 TON, TEST 3 

FAILURE, TEST 3 



P345 



SECTION F-F 
P342 



P344 



P343 




ED 



SECTION I -I 



£L 



P503 



P504 



SECTION K-K 
P502 P50I 




FIGURE 2.32. HORIZONTAL VARIATION IN PORE PRESSURE DURING REFERENCE 
PILE TESTS (1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ; 1 ton = 8.9 kN) 



99 



9- 



x 

UJ 

Q 



41- 



50-J 



2- 



SECTION B-B £ 5 - 




PI95 



PI93 



15- 
10 

5 





P5 



SECTION E-E 

P2 



P3 



PA 




P345 



SECTION H-H 



(P343) 



35r 

CO 3° 

Q- 251- 

Z 20 

~ 15 

Z> 
CO 
V) 
UJ 

or 
a. 





P5 




P3 


P4 


- 






^r^^^^ 








4 k 

11 


- 

















P5 



SECTION J-J 
P2 



P3 



P4 



30p 
25- 
20- 
15- 
10- 
5- 
OL- 




SECTION K-K 



40 
35 
30 
25 
20 
15 
10 
5 
O 



P503 


P504 


P502 


P50I 


I 


^ 


*& 


K 








LEGEND 


c 


f 


7 


N 


\ 


F"" 


^ 


• 300 TON, TEST 

j o- -FAILURE, TEST 

► A 300 TON, TEST 

A FAILURE, TEST 

■ 300 TON, TEST 

D FAILURE, TEST 



2 
2 
3 
3 



FIGURE 2.33. HORIZONTAL VARIATION IN PORE PRESSURE DURING 9-PILE TESTS 
(1 ft = 0.305 m; 1 psi =6.89 kN/m 2 ; 1 ton =8.9 kN) 



100 



9- 



19- 



~ 34- 



Q. 

UJ 

o 



41- 



50-1 




SECTION E-E 



PI95 PI93 




SECTION H-H 



P345 



P503 



SECTION K-K 
P504 P502 



P50I 




P344 P342 (P343) 

P4_ 

DA« 




LEGEND 

■ 300T0N, TEST 3 

□ FAILURE, TEST 3 

A 200T0N, 5-PILE TEST 

A FAILURE, 5-PILE TEST 

• 200T0N.4-PILE TEST 

O FAILURE, 4-PILE TEST 



FIGURE 2.34. HORIZONTAL VARIATION IN PORE PRESSURE DURING SUBGROUP TESTS 
(1 ft = 0.305 m; 1 psi = 6.89 kN/m 2 ; 1 ton =8.9 kN) 



101 



pore pressure conditions in and around the group before the 9-pile 
tests; and Fig. 2.31 pertains to the conditions preceding the subgroup 
tests. Pore pressure variations for the third (final) 9-pile group test 
are also included in Fig. 2.31 for purposes of comparison. 

A fairly significant scatter in the absolute values of pore pressure 
can be seen from point to point in these figures. Analyzed in total, 
however, the average pore water pressures on the piles and in the soil 
at the two common depths, 19 ft. and 34 ft. (5.8 m and 10.4 m), at 
which pile and ground piezometers were placed did not differ 
significantly prior to the tests, which is a further indication that excess 
pore pressures generated during driving had almost fully dissipated 
through-out the entire mass of soil controlling the behavior of the 
group. The pressure gradients implied in the various figures under 
consideration probably do not have much physical significance. 

Figures 2.32-2.34 show measured pore pressures at about one-half 
the failure load and at failure for the reference pile tests, 9-pile group 
tests, and subgroup tests, respectively. The pore water pressure 
changes in the soil mass, as on the faces of the piles, were very small 
during both reference pile and group tests. At some soil instrument 
locations the changes were negative, while at others the changes were 
positive, but almost all changes recorded in the soil mass, as on the 
piles, from prior to a test until failure were less than about 2 psi (14 
kN/m 2 ). Pore pressure changes during load test in the soil 
surrounding Pile 1 were not significantly different from those in the soil 
mass surrounding the group piles. 

Depthwise Variations of Lateral Pressures on Piles . The observed 
vertical variation in pore water pressure on the face of Pile 1, the 
reference pile, prior to each compression load test and at failure is 
plotted in Fig. 2.35. That figure, as well as the following figures on 
vertical pore pressure variation, also shows a "hydrostatic line" that is 
a graph of pore pressure variation based on a constant piezometric 
surface at a depth of 7.5 ft. (2.3 m), which was the average free 
water surface depth in the soil borings as well as in the anchor casings. 
Some trend toward slightly lower pressures with each succeeding test 
can be observed, and the small pressure changes developed during 
loading, described previously, are evident. The low pore pressure 
value at the third level is believed to be unrepresentative. 

The average vertical pore water pressure variation with depth for 
the four instrumented group piles is shown in Figs. 2.36 and 2.37. 
These figures represent the 9-pile and subgroup compression tests, 
respectively. On the latter figure, the pore pressure variation for the 
third 9-pile test is shown for purposes of comparison. Average results 
for the three tests reported in Fig.. 2.37 consider only the piles 
actually being loaded. Hence, different numbers of piezometers are 
included in each average, so that conclusions concerning apparent 
profile changes between the 9-pile group and subgroup tests should not 



102 



WATER LEVEL 
IN BORINGS 



43 
FIGURE 2.35 




PORE WATER PRESSURE (PSF) 
2000 3000 4000 5000 



i 



PILE NUMBER I 
(REFERENCE PILE) 



LEGEND 



D PRETEST PORE PRESSURE 
■ PORE PRESSURE AT FAILURE 

TEST I 

TEST 2 

TEST3 



HYDROSTATIC LINE 



VERTICAL VARIATION IN PORE PRESSURE ON REFERENCE PILE 1 
(1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



103 



43*- 



1000 

r~ 



WATER LEVEL 
IN BORINGS 




PORE WATER PRESSURE ( PSF) 
2000 3000 4000 



5000 
» 



PILE GROUP AVERAGE 



LEGEND 

D PRETEST PORE PRESSURE 

■ PORE PRESSURE AT FAILURE 

9 -PILE TEST I 

9- PILE TEST 2 

9- PILE TEST 3 






HYDROSTATIC LINE 



FIGURE 2.36. VERTICAL VARIATION IN AVERAGE PORE PRESSURE ON GROUP PILES; 
9-PILE TESTS (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



104 



PORE WATER PRESSURE (PSF) 
IOOO 2000 3000 



T 



T 



T 



4000 
1 — 



5000 
r 



5- 



P1LE GROUP AVERAGE 



WATER LEVEL 
IN BORINGS 




LEGEND 



□ PRETEST PORE PRESSURE 
■ PORE PRESSURE AT FAILURE 
9-PILETEST 3 (PILES 2,3,4,5 



5 -PILE SUBGROUP TEST (PILES 2,3,5)° 

4-PILE SUBGROUP TEST (PILES 3,5)° 

a- NOT A COMPLETE AVERAGE 
TO BE COMPARED WITH 
TEST 3 



HYDROSTATIC LINE 



VERTICAL VARIATION IN AVERAGE PORE PRESSURE ON GROUP PILES- 
SUBGROUP TESTS (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



105 



be drawn. It is evident, however, that the vertical variation of pore 
pressure against the group piles was essentially hydrostatic, both prior 
to each group test and at failure. Some slight trend toward reduction 
in pore pressure between 9-pile tests can be observed in Fig. 2.36. 

Figure 2.38 shows the vertical pore pressure variation on the four 
piles subjected to individual uplift tests that were instrumented for 
lateral pressure measurement. Again, for the uplift tests, pressure 
variation was essentially hydrostatic, with only small changes developing 
during loading. On all piles pore pressure changes were slightly 
positive at the upper two levels and essentially zero at the lower two 
levels during uplift loading. 

Graphs of total pressure variation with depth, corresponding to 
the graphs of pore water pressure variation with depth that were shown 
in Figs. 2.35-2.38, are presented in Figs. 2.39-2.42. The observed 
patterns in the reference pile (Fig. 2.39) and for the average of the 
group piles (Fig. 2.40) are discernably different, with the total 
stresses at the first three levels being much higher in the group piles 
than in the reference pile. This difference is attributed to the effects 
of data scatter, discussed elsewhere. Pressures on the group piles are 
probably more reliable than those shown for the reference pile because 
each point is the average of readings on several piles. 

Indicated total pressures decreased slightly, but relatively 
uniformly, at all depths on the group piles between Nov. 5, four days 
after completion of driving, and Nov. 16, the date of the first load 
test. No such behavior was observed for Pile 1. This effect could not 
be attributed to any operational problem but may be related to 
temperature correction methods. After Test No. 1, there was a general 
increase in indicated total stress in the upper level in the reference 
pile and in the upper three levels in the group piles. The evidence 
compiled through analysis of maximum side load transfer in the various 
tests (Appendix D) does not support an increase in total (and 
consequently effective) lateral normal stresses of the magnitude 
indicated in Figs. 2.39 and 2.40. Therefore, the total stress values in 
Figs. 2.39 and 2.40 should be considered as general trends and not 
precise representations of the states of total stress. This statement 
also applies to the calcualted effective stress variations shown later. 

Total lateral pressure variation on the group piles in the subgroup 
tests approximated those in the third 9-pile test, as shown in Fig. 

2.41. Total pressure profiles for the uplift tests aer shown in Fig. 

2.42. Note that the total stress variation in uplift for the piles that 
were in the group does not include data from Pile 3, which was not 
subjected to an uplift test. The absence of data for that pile caused 
the average indicated total pressures for the uplift tests to differ 
significantly from those for the earlier compression tests on the group 
at the upper two levels. 



106 



PORE WATER PRESSURE (PSF) 

IOOO 2000 

1 — 



UPLIFT TESTS 



LEGEND 



PRETEST PORE PRESSURE 

PORE PRESSURE AT 
FAILURE 



REFERENCE PILE 
NO. 



AVERAGE OF GROUP 

PILES 2,4,5 




3000 



HYDROSTATIC 
LINE 



FIGURE 2.38. VERTICAL VARIATION IN PORE PRESSURE ON PILES DURING 
UPLIFT TESTS (1 ft = 0.305 m; 1 psf =47.9 N/m 2 ) 



107 







TOTAL LATERAL PRESSURE (PSF) 
5000 10,000 



15,000 



NUMBER 



(REFERENCE PILE) 



LEGEND 



AFTER DRIVING (NOV. 5, 1979) 
PRETEST TOTAL PRESSURE 
TOTAL PRESSURE AT FAILURE 
TEST NUMBER I 
TEST NUMBER 2 
TEST NUMBER 3 







VERTICAL VARIATION IN TOTAL LATERAL PRESSURE ON REFERENCE 
PILE 1 (I ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



108 



TOTAL LATERAL PRESSURE (PSF) 
5000 10,000 



15,000 



10 



15 



t 
-20 

Q_ 
Ixl 
Q 

25 



30 - 



35 



40 



43 L 



1— 1 ■ 1 

PILE GROUP AVERAGE 
LEGEND 

TEST NUMBER I X AFTER DRIVING (NOV. 5, 1979) 

TEST NUMBER 2 O PRETEST TOTAL PRESSURE 

TEST NUMBER 3 • TOTAL PRESSURE AT FAILURE 




FIGURE 2.40. VERTICAL VARIATION IN AVERAGE TOTAL LATERAL PRESSURE ON 
GROUP PILES (2, 3, 4, 5) (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



109 



TOTAL LATERAL PRESSURE (PSF) 
5000 10,000 



15,000 



I 5-- 



Li. 

r 20 



Q. 
Ill 
Q 



25- 



30- 



35- 



40- 



43 L 



PILE GROUP AVERAGE 
LEGEND 

O PRETEST TOTAL PRESSURE 
• TOTAL PRESSURE AT FAILURE 

TEST NUMBER 3 

• 5-PILE SUBGROUP (a) 

— 4-PILE SUBGROUP (a) 



(a) INCLUDES RDGS. ON 
UNLOADED PILES 




FIGURE 2.41. VERTICAL VARIATION IN AVERAGE TOTAL PRESSURE ON PILES IN 
SUBGROUP TESTS (2, 3, 5) (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 

110 



10 - 



5 - 



- 20 r 

x 



Q_ 

l±J 
Q 



TOTAL LATERAL PRESSURE (PSF) 
5000 10,000 15,000 



20,000 



T 



25 - 



30- 



35 



40 - 
43- 

FIGURE 2.42 



I 1 

LEGEND 

PRETEST TOTAL PRESSURE 

TOTAL PRESSURE AT FAILURE 

UPLIFT TEST NUMBER I (REFERENCE PILE I) 

PILE GROUP AVERAGE FOR UPLIFT TESTS (a) 

(a) PILES 2,4,5 ONLY 




VERTICAL VARIATION IN TOTAL PRESSURE FOR UPLIFT TESTS 
(1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 

111 



Figures 2.39-2.42 clearly demonstrate, in profile form, that total 
pressure changes generated on the pile surfaces due to loading were 
uniformly small. 

Finally, normal effective stresses against the faces of the piles 
were calculated by subtracting the pore water pressure readings 
detailed in Figs. 2.35-2.38 from the total pressure values given in Figs. 
2.39-2.42. The resulting effective stress variations, for pretest and 
failure conditions for all tests, are given in Figs. 2.43-2.46 in a format 
corresponding to the pore and total stress variations described 
previously. The patterns of effective stress are dominated by the 
observed total stress patterns, and the combination of the small 
measured pore and total pressure changes during loading result in small 
normal effective pressure changes. 

Ground Movements During Tests 

Vertical soil movements at several points on the surface of the soil 
and at several depths were monitored during the load tests to provide a 
means of verifying the mode of failure (punching of individual piles 
versus failure of the group as a block), to obtain information on the 
lateral extent of the zone of surface soil deformations produced by 
loading the pile groups, and to obtain deformation data that could be 
used by others to calibrate mathematical models which calculate soil 
deformations (e.g., the finite element model). 

Some irregulrities due to ambient temperature changes, "bumping" 
of gages, and similar effects, were observed when the raw ground 
settlement readings were plotted as functions of time. Plots of raw 
ground movement data versus applied load for the 5-minute reading set 
are contanied in Appendix F. The raw data plots for each settlement 
point were first smoothed by a procedure outlined in Appendix F to 
correct for false readings produced by rapid ambient temperature 
changes. A separate study, documented in Appendix E, was conducted 
to evaluate the expected range of false readings to be expected from 
temperature fluctuations, and this information was used in correcting 
the data. 

Several other considerations in interpreting the ground movement 
data included (1) discarding the data for the first reference pile test 
due to inadequate shading of the dial gages stands; (2) discarding the 
data for the subgroup tests for the 600 inch (15.2 m) depth due to 
unexplained inconsistencies; and (3) combining and averaging the data 
for each settlement point in groupings according to reference tests, 
9-pile group tests, and subgroup tests without considering individual 
tests. Analysis of the raw data revealed that very minor differences 
existed between the soil deformations in the first and third test sets for 
both the reference and group piles and between the two subgroup tests 
and that measured settlements were essentially recoverable. 



112 







EFFECTIVE LATERAL PRESSURE (PSF) 
5000 10,000 



5,000 



T 



(REFERENCE PILE) 
LEGEND 




NUMBER 



AFTER DRIVING (NOV. 5, 1979) 
PRETEST EFFECTIVE PRESSURE 
EFFECTIVE PRESSURE AT FAILURE 
TEST NUMBER I 
TEST NUMBER 2 
TEST NUMBER 3 



VERTICAL VARIATION IN LATERAL EFFECTIVE STRESS ON REFERENCE 
PILE 1 (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



113 



EFFECTIVE LATERAL PRESSURE (PSF) 

5000 10,000 15,000 



15 



- 20 

x 
h- 

Q_ 
LU 

° 25 



30 

35 

40 
43 



PILE GROUP AVERAGE 
LEGEND 

TEST NUMBER I X AFTER DRIVING (NOV.5,1979) 

TEST NUMBER 2 A PRETEST EFFECTIVE PRESSURE 

TEST NUMBER 3 A EFFECTIVE PRESSURE AT FAILURE 




FIGURE 2.44. VERTICAL VARIATION IN AVERAGE LATERAL EFFECTIVE STRESS ON 
GROUP PILES (2, 3, 4, 5) (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



114 



EFFECTIVE LATERAL PRESSURE (PSF) 

5000 10,000 15,000 



Q_ 
UJ 
Q 



15 



Ll. 

- 20 k 

x 



PILE GROUP AVERAGE 
LEGEND 

A PRETEST EFFECTIVE PRESSURE 
▲ EFFECTIVE PRESSURE AT 
FAILURE 

. TEST NUMBER 3 

5- PILE SUBGROUP 

4- PILE SUBGROUP 




25- 



30- 



35- 



40- 



43 L 

FIGURE 2.45. VERTICAL VARIATION IN AVERAGE LATERAL EFFECTIVE STRESS IN 
SUBGROUP TESTS (PILES 2, 3, 5) (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



115 







10 - 



- 20 

x 

H 

a. 

IxJ 

°25 



30 



EFFECTIVE LATERAL PRESSURE (PSF) 

5000 10,000 15,000 



A 

A 



35 

40 
43 

FIGURE 2.46. 



LEGEND 

PRETEST EFFECTIVE PRESSURE 

EFFECTIVE PRESSURE AT FAILURE 

UPLIFT TEST NUMBER I (REFERENCE PILE I) 

PILE GROUP AVERAGE FOR UPLIFT TESTS (a) 



(a) PILES 2,4,5 ONLY 




VERTICAL VARIATION IN LATERAL EFFECTIVE STRESS FOR UPLIFT 
TESTS (1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



116 



The corrected and averaged readings are presented in graphical 
form in Figs. 2.47-2.54. In these figures the settlement points are 
arranged in order according to their distance from either the reference 
pile (Pile 1 ) or the center of the pile group, which are assumed to be 
points of symmetry. This arrangement, considered without regard to 
direction of the settlement point from the assumed point of symmetry, 
may be the cause of the reading variations that are seen in these 
figures. Assessment of errors by the authors indicate a reliability of 
about ± 0.005 in. (0.13 mm) in the plotted results. 

Figure 2.47 shows soil surface displacements for the last two 
reference pile tests at a distance interval of about 40 to 80 in. (1.0 to 
2.0 m) from the center of the pile, which had a radius of 5.375 in. 
(137 mm), and pile displacements. The distance from the center to the 
edge of the pile is represented by the shaded zone on the figure. The 
displacements measured in the reference tests (Fig. 2.47) indicate (1) 
that virtually all of the surface soil deformation occurred within less 
than one meter (about 4 diameters) of the face of the pile, and (2) that 
no heave or otherwise unusual soil movements occurred at failure. 

Figure 2.48 depicts the surface soil movements in the soil mass in 
and around the 9-pile group. Several facts are evident in this figure: 
(1) failure was of the punching type, i.e., the soil near the piles, both 
inside and outside the group, did not move down with the piles as the 
piles failed; (2) soil-pile deformations were essentially horizontally 
continuous up to an applied load of about 400 kips (1780 kN); and (3) 
no surface heave accompanied failure. Soil surface movements at failure 
did not exceed about 0.02 in. beyond a distance of 170 in. (4.3 m) 
from the center of Pile 2. The predominant zone of soil straining was 
observed to be within about 3 in. (75 mm) of the faces of the piles. 

Similar soil surface deflections are seen for the average of the two 
subgroup tests in Fig. 2.49. However, soil settlements in the vicinity 
of the piles can be seen to be only about one-half of those generated in 
the 9-pile tests at equivalent loads. Lateral gradients of deflection are 
also notably lower. 

Movements at two locations in the soil at depths of 300 in., 516 
in., and 600 in. (25 ft., 43 ft., and 50 ft.) (7.6 m, 13.1 m, and 15.3 
m) for the 9-pile group and for the average of the subgroups are 
shown in Figs. 2.50-2.54. (No graph is shown for the lowest level for 
the subgroup tests, as no reliable data were acquired.) The trends in 
subsurface deflections are very similar to the trends in surface 
deflections. Examination of the data indicates that the average shear 
strains in the soil between DSP1 and DSP2, situated between about 10 
and 60 inches (250 and 1520 mm), respectively, from the face of 
edge Pile 9, were in the order of 0.04 per cent at and above a depth of 
25 ft. (7.6 m) and in the order of 0.01 percent at the 43 and 50 ft. 
(13.1 and 15.3 m) depths for a nominal load value of 800 kips (3560 
kN) in the 9-pile group. This load would probably be somewhat in 



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NOMINAL SINGLE PILE LOAD 

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NOTE : SSP3 WAS ON 
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REFERENCE BEAM SUPPORT 



m NOTE : ALL SOIL SETTLEMENT READINGS WERE 
ZEROED AT BEGINNING OF EACH TEST 



FIGURE 2.47. SURFACE SOIL MOVEMENTS NEAR PILE 1; REFERENCE TESTS 
(1 lb = 4.45 N; 1 in = 25.4 nun) 



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OPPOSITE (SOUTH) SIDE 
FROM REMAINING POINTS. 



FIGURE 2.48. SURFACE SOIL MOVEMENTS FOR 9-PILE GROUP TESTS 
(1 lb = 4.45 N; 1 in = 25.4 mm) 



119 



HORIZONTAL DISTANCE FROM CENTER OF PILE CAP (IN.) 



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HAD HIGHER RESISTANCE 
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FIGURE 2.49. SURFACE SOIL MOVEMENTS FOR AVERAGE OF SUBGROUP TESTS 
(1 lb = 4.45 N; 1 in = 25.4 mm) 



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2 AND 3) 



FIGURE 2.50. SOIL MOVEMENTS; 300 INCH (7.6 M) DEPTH: 9-PILE GROUP 
TESTS (1 lb = 4.45 N, 1 in = 25.4 ram) 



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FIGURE 2.51. SOIL MOVEMENTS; 300 INCH (7.6 M) DEPTH; AVERAGE OF 
SUBGROUP TESTS (1 lb .= 4.45 N; 1 in = 25.4 mm) 



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FIGURE 2.52. SOIL MOVEMENTS; 516 INCH (13.1 M) DEPTH; 9-PILE GROUP 
TESTS (1 lb = 4.45 N; 1 in = 25.4 mm) 



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FIGURE 2.53. SOIL MOVEMENTS; 516 INCH (13.1 M) DEPTH; AVERAGE OF 
SUBGROUP TESTS (1 lb = 4.45 N; 1 in = 25.4 mm) 



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NOMINAL GROUP LOAD 



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FIGURE 2.54. SOIL MOVEMENTS; 600 INCH (15.3 M) DEPTH; 9-PILE GROUP 
TESTS (1 lb = 4.45 N; 1 in = 25.4 mm) 



125 



excess of a working load for this group. These strain values are based 
on the assumption that all soil movement is vertical in the vicinities of 
settlement monuments, and, while such an assumption may not be 
completely correct, particulary near the pile tips, the quoted shear 
strain magnitudes suggest that the elastic modulus to be selected for 
pile-soil-pile interaction calculations in the hybrid model should be 
taken at very low strain amplitudes, preferably at less than 0.1% 
principal normal strain in heavily overconsolidated clays of the type 
encountered at this test site if triaxial or pressuremeter tests are used 
to evaluate soil def ormability . 



126 



Chapter 3. Load Transfer 

General 

The objectives of this chapter are to present detailed information 
on measured load transfer, to compare load transfer patterns for 
reference and group piles, to assess the effects of residual stress on 
load transfer, and to correlate load transfer with measured soil 
properties. With regard to these objectives, comments are made con- 
cerning certain basic phenomena thought by the authors to be involved. 

Load Transfer Patterns for Reference Piles and for Group Piles by 
Position 

The average disturbutions of load along the reference piles, center 
group pile, edge group piles, and corner group piles are plotted in 
Figs. 3.1 - 3.10 for the three 9-pile test sets, for the 5-pile subgroup 
test, and for the 4-pile subgroup test. Comparative graphs are 
presented for a load value in the working load range and for the 
average peak failure condition. These figures, which do not include 
the effects of residual loads (discussed later) or cap weight, also show 
the corresponding relationships of developed unit side shear (f) to 
depth (d), which were obtained by differentiation of the load distri- 
bution curves assuming that the shearing surface coincided with the 
pile surface. 

The load distribution curves were derived from the raw load data 
using a piecemeal second degree least-squares fitting procedure, 
described in Appendix E, which also gives an example of the differ- 
ences between raw and fitted load distributions. In general these 
differences were very small except at the points in the f-d curves 
where the greatest curvature exists, where the fitted f values were 
somewhat smaller than the raw f values in the 20 ft. (6.1 m) depth 
range and larger in the 28 ft. (8.5 m) depth range. 

While only average values of load and unit side shear have been 
plotted in Figs. 3.1 - 3.10, a sense of the scatter of the data can be 
obtained by consulting the load transfer corelation tables given later in 
this chapter. Appendix D presents an extensive set of load- settlement, 
load-distribution, f-d, and f-z (unit load transfer vs. relative 
deformation curves) for all tests beyond 9-pile Set 1, which is covered 
in this chapter. 

Reference to Figs. 3.1, 3.3, and 3.5 reveals that the group piles 
developed substantially different load transfer patterns than occurred in 
the reference piles at an average load per pile of about 60 kips (267 
kN). The reference piles, at that load, developed considerably higher 
rates of load transfer in Zones A and B (Fig. 1.1) than did any of the 
group piles, while greater side load transfer was registered in the 
group piles in Zone D and in end bearing. Observation of the f-d 



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curves shows a clear trend toward decreasing load transfer in Zone B 
with increasing "protection" of the pile. That is, the center pile 
transferred less load than the edge piles, which in turn transferred 
less load than the corner piles. 

The trend is not so clear among the group piles near the bottoms 
of the piles, but the group piles as a whole clearly transferred more 
load there than did the reference piles simply because of the lack of 
ability to transfer load at a higher level. 

The above phenomena have been observed by others in both full 
scale and model tests and are believed to be a general characteristic of 
pile group behavior, although the extent to which differences between 
load transfer in isolated and group piles exists is expected to be a 
function of relative pile-soil flexibility. The trend toward load transfer 
occurring farther down the piles at subfailure loads for piles in groups 
and to occur at the greatest depth for the most protected (interior) 
piles is due to the decreased opportunity for relative deformation to 
occur between the group piles and soil near the surface because near- 
surface soils have been forced to settle to a greater extent than the 
corresponding soils around an isolated pile due to stress overlaps 
produced by the group piles. 

It should be observed that essentially no load was transferred in 
the upper 5 ft (1.5 m) of any of the piles tested. This was within the 
depth of predrilling but was also in the zone of highest OCR and 
greatest lateral pile motion, both during driving and testing. 

Other aspects of the shapes of the subfailure f-d curves can be 
explained qualitatively in terms of measured soil properties. The peaks 
observed at a depth of approximately 20 ft. (6.1 m) correspond to a 
zone of high lateral in-situ pressures (Fig. 1.4), while the reduced 
load transfer in the vicinity of 28 ft. (8.5 m) is associated with Zone C 
(Fig. 1.1), which is a softer (and probably less overconsolidated) soil 
than that in the zones above and below. The high load transfer in Soil 
Zone D is most probably associated with depthwise increasing sand 
content in the sandy clay comprising that zone, depthwise increasing 
in-situ lateral earth pressures, and the low volumetric compressibility of 
the soil in Zone D, as expressed in Fig. 5.6 of the Interim Report . 

The patterns of load transfer in the subgroup tests at loads 
approximately equal to one-half of failure load, shown in Figs. 3.7 and 
3.9, reveal, as expected, somewhat smaller differences between 
reference pile and group pile behavior, especially in the 4-pile test 
where loaded pile spacing was 4.2 diameters. In fact, the differences 
appear as large as they are in these figures principally because 
comparisions had to be made at significantly different pile head loads 
due to the fact that coincidental reference and group pile-head loads 
were not achieved in the subgroup tests. 



138 



Figures 3.7 and 3.9 also depict the patterns of load transfer 
induced in the unloaded piles during the subgroup tests. The 
developed unit side shear patterns resembled side shear patterns for 
piles undergoing downdrag, with negative side resistance (downdrag) 
above about 30 ft. (9.2 m) and positive side resistance below that 
depth. The zone of maximum negative load transfer occurred about 1 m 
below the depth of maximum positive load transfer in the loaded piles. 
Only minor differences in load transfer in the original center and corner 
piles existed at this load level in the 4-pile subgroup test. Depth wise 
maximum negative side shear stresses induced in the unloaded piles for 
the configuration shown in the figures were about 20 percent of the 
depthwise maximum positive side shear stresses in the loaded group 
piles in the 4-pile test and about 25 percent in the 5-pile test. 

Distributions of loads along piles at failure and values of f versus 
d at peak load are described in Figs. 3.2, 3.4 and 3.6 for the 9-pile 
group tests and in Figs. 3.8 and 3.10 for the subgroup tests. These 
load transfer curves were derived by averaging the load distributions 
in the indicated piles (e.g., corner group piles) at the maximum pile- 
head load developed for each pile. Such maximum loads were not 
always developed simultaneously on all piles in a set. The curves also 
do not represent the absolute maximum value of f achieved at every 
level because side resistance failure was distinctly progressive, as will 
be described subsequently. They do represent the available side 
resistance at failure, however, which is the quantity of interest to 
designers. 

A more uniform f-d pattern existed among the various group and 
reference piles at failure than existed at 60 kips (267 kN) per pile. 
This is especially true for the first 9-pile load test, depicted in Fig. 
3.2. An increasing dissimilarity of load transfer pattern occurred with 
further testing, especially in Zone B, possibly because of the degrading 
effects of multiple loading in this soil. 

The shapes of the f-d curves at failure were similar to those at 
the 60-kip (267 kN) load level, except* for the sharp increase in unit 
side resistance near the bottoms of the piles. High unit side shear did 
not exist in the lower portions of the piles at the lower loads because 
relative pile- soil movement had not yet occurred that was sufficient to 
develop as high a percentage of maximum unit side resistance at that 
level as was developed farther up the piles, especially in the reference 
piles. This phenomenon, well-known from tests on instrumented single 
piles, is associated with the compression that takes place within flexible 
piles, which results in larger downward movements in the piles near 
the tops than near the tips for a given applied load. 

The depths of median side load transfer, tabulated in Table 3.1, 
were smaller in the reference piles at subfailure loads than were the 
corresponding depths for the group piles. At failure, however, all 
median load transfer depths were essentially equal at slightly above the 
two-thirds depth. 

139 



TABLE 3.1. VARIATION OF DEPTH OF MEDIAN SIDE LOAD TRANSFER AMONG TESTS 

(1 ft = 0.305 m; 1 k = 4.45 kN) 

DEPTH OF MEDIAN SIDE LOAD TRANSFER (FT) 



TEST 


LOAD 
LEVEL 


AVG. OF 

REFERENCE 

PILES 


CENTER 
PILE 


AVG. OF 
EDGE PILES 


AVG. OF 
CORNER PILES 


9 -PILE 
TEST 1 


60K/PILE 
FAILURE 


21 
26 


26 
27 


27 
25 


26 
26 


9- PILE 
TEST 2 


60K/PILE 
FAILURE 


21 
25 


25 
28 


25 
26 


23 
27 


9- PILE 
TEST 3 


60K/PILE 
FAILURE 


20 
24 


25 
26 


25 
27 


24 
26 


5- PILE 
SUBGROUP 


60K/PILE 
FAILURE 





25 
29 


19 
26 


— 


4- PILE 
SUBGROUP 


60K/PILE 
FAILURE 





— 


20 
24 






NOTE : MIDDEPTH OF PILES = 21.5 FT. 

2/3 OF PILE PENETRATION = 28.7 FT. 



140 



No discernable "tip effect" was observed for either the group piles 
or the reference piles at loads up to and including those producing 
shaft failure or total failure, which occurred simultaneously except for 
reduction in unit load transfer near the tip of a pile caused by the 
influence of the stress field generated in the soil at the pile tip on the 
stress field in the zone around the pile just above the tip. In this 
regard it should be emphasized that the lowest two load transducer 
levels were situated 1 ft. (0.305 m) and 4 ft. (1.22 m), respectively, 
above the bottoms of the boot plates. Therefore, the center of the 
lowest increment of pile load measurement was 2.5 ft. (0.76 m) or 2.8 
diameters above the pile tips, which should be considered the lowest 
level at which the f-d data are applicable, so that some undetected tip 
effect may have existed before or at shaft failure. Tip movements were 
very small at failure. However, further tip penetration was accompanied 
by relaxation of load transfer in the bottom few feet of the piles. This 
may have been a manifestation of tip effect. 

In the subgroup tests , the magnitudes of the side shear stresses 
induced in the unloaded piles at failure were greater than those report- 
ed for the 60 kip (267 kN) per pile load condition. Transition from 
negative to positive load transfer also occurred at a greater depth at 
failure than at the lower load. This resulted in a corresponding higher 
induced tip load in the unloaded piles. Consideration of the induced 
load transfer (f-d) patterns provides a clear phenomenological explana- 
tion of the reasons for the differences in load transfer patterns in 
single piles and in piles within groups. 

Apparent Peak Load Transfer by Soil Layer 

If the average peak load transfer values depicted in Figs. 3.2, 
3.4, 3.6, 3.8, and 3.10 are tabulated by soil zone or layer, Table 3.2 
results. The word "apparent" is used here because the side shear 
stresses tabulated are only the stresses mobilized upon application of 
external load and do not consider side shear stresses that existed due 
to residual loads in the piles. It is significant to observe that all 
apparent peak load transfer values (denoted f max ) below a depth of 11 

ft. (3.4 m) exceed 1 ksf (47.9 kN/m 2 ), a value sometimes considered as 
a limiting value for skin friction in overconsolidated clay. In fact, 

f exceeded 1.5 ksf (71.9 kN/m 2 ) in Zone D. There was also a 

trend toward increasing values of f max i n eacn layer with repeated 

compression loading, except in Pile 2, the interior pile, where a trend 
toward side shear degradation is evident in Zones B and C. 

The only significant differences in apparent peak load transfer 
between the reference and group piles occurred in Layers A and B. In 
both layers the reference piles produced slightly higher apparent load 
transfer values. An anticipated large increase in the load transfer 
value in the group piles in the zone of predrilling did not materalize, 
indicating that installation of nearby piles had no discernable effect on 

141 



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radial lateral pressures around the upper parts of piles driven in 
shallow pilot holes. 

Progressive Failure Patterns 

The soil at the test site can be described as "brittle" and "strain 
softening." (Refer to Appendix C for typical laboratory stress-strain 
relationships.) Because of this fact pile failure was progressive in two 
ways: (1) progressive failure occurred along the shaft of each pile, 
and (2) progressive failure occurred among the piles of the group. 
After the soil near the bottoms of the piles failed, some relaxation 
occurred, possibly due to the tip effect described earlier. As a result 
of this phenomenon and of the small relative deformations needed to 
mobilize peak load transfer in Zone D, failure often progressed from the 
bottoms of the piles toward the tops at the same time failure was 
progressing from the tops toward the bottoms. This phenomenon is 
presented diagramatically in Figs. 3.11 and 3.12, which show by means 
of vertical black bars the range over which shaft failure had progressed 
in each pile as a function of applied load for Test Set 1. No shear 
failure was observed anywhere along the shafts at loads lower than 
those described. All piles, except for Piles 3 and 11, show upward 
progression of failure from near the pile tips beginning at loads below 
the failure load. It is suggested that the load at which depthwise 
progressive failure begins would be the approximate load at which the 
group or reference piles would have failed under long-term sustained 
loading. 

Figure 3.12 also shows how shaft failure (and total failure, which 
corresponded to complete shaft failure) progressed from pile to pile 
during the first 9-pile test, in which tipping of the cap toward the 
north row of piles (8, 9, and 10), along with some rotation about a 
vertical axis and northward, translation, occurred. See Fig. 2.8. Pile 
8, the northeast corner pile, failed first, followed by Piles 9 and 10 
and the piles in the middle row. Pile 6, shown as having incomplete 
shaft failure in Fig. 3.12, did fail for all practical purposes during the 
one-hour load hold at 700 tons (6.23 mN). 

Effects of Residual Stresses on Load Transfer 

Measured variations in the residual loads within the reference piles 
before and after Test 1 are shown in Fig. 3.13. These loads are based 
on zeros taken while the piles were in an unstressed condition in the 
calibration beds. It can be observed that the residual loads induced by 
load testing the piles to failure exceed by a significant amount the 
residual loads induced during driving. This residual side shear stress 
pattern was computed by differentiation of the graph of residual load 
before Test 1 versus depth. 

Figure 3.14 shows the average residual side shear stress distribution 
corresponding to the residual load distribution patterns observed prior 
to Test 1. Negative residual stresses (downward directed on piles) of 

143 



(GROUND 

SURFACE) 3 4 5 




u. 20- 



UJ 
O 



40- 



n 




INDICATES LOAD 
NUMBER * 



*3=40T NOMINAL APPLIED 
4= 50T NOMINAL APPLIED 
5=65T NOMINAL APPLIED 
(PRIOR TO PLUNGING) 



* 8= 80T NOMINAL APPLIED 
9=85T NOMINAL APPLIED 
10= 80T NOMINAL APPLIED 
(DURING PLUNGING) 



FIGURE 3.11. 



PROGRESSIVE FAILURE IN REFERENCE PILES; TEST 1 
(1 ft = 0.305 m; 1 ton = 8.9 kN) 



144 



(GROUND 
SURFACE) 5 6 




y 



INDICATES LOAD 
NUMBER FOR TEST r 

56 7 8 





5 6 7 8 



1 



PILE 10 



PILE 9 



5 6 7 8 



5 6 7 8 



20- 




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j 



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J 



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6=600T NOMINAL APPLIED 

7 = 650T NOMINAL APPLIED 

8 = 700T NOMINAL APPLIED 

FIGURE 3.12. PROGRESSIVE FAILURE IN GROUP PILES; 9-PILE TEST 1 
(1 ft = 0.305 m; 1 ton = 8.9 kN) 



145 



RESIDUAL LOAD CK) 
10 20 30 40 



UPPER BOUND CPILE I) 




/ 31' LEVEL 
/ INOPERATIVE 
/ ON PILE I 



43 L 



FIGURE 3.13. RESIDUAL LOADS IN REFERENCE PILES; TEST 1 
(1 ft = 0.305 m; 1 k = 4.45 kN) 



146 



f CPSFD 
500 1000 



2000 




/-AT MAX SIDE SHEAR 
/ STRESS: WITHOUT 
\ CONSIDERING RESIDUAL 

\ STRESS EFFECTS 



AT FAILURE'- 
WITHOUT 
CONSIDERING 
RESIDUAL STRESS 
EFFECTS 



AT MAX SIDE 
SHEAR STRESS: 
n^ CONSIDERING 
k \ RESIDUAL 
STRESS 
EFFECTS 



43 L 



FIGURE 3.14. F-D RELATIONSHIPS FOR REFERENCE PILES; TEST 1 
(1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



147 



up to 400 psf (19 kN/m 2 ) were observed in the top two-thirds of the 
piles, while lesser positive values existed in the bottom one-third. 

Figure 3.14 also shows graphs of maximum apparent side shear 
distributions (without considering group effects) based on the average 
side shear stresses in Piles 1 and 11 in Test 1 at the time of total pile 
failure. It also shows curves based on the average maximum value of 
measured side shear stresses at each level, irrespective of whether 
those values occurred at the time of total failure. The average measured 
initial residual side shear stress distribution was then added to these 
apparent f-d curves to produce "true" f-d curves. The true f-d 
curves represent the actual variation of peak stress with depth 
experienced by the soil at the time the reference piles failed and the 
variation of maximum peak shear stress, respectively. It can be seen 
that consideration of residual stresses tends to linearize the graphs of 
f versus depth, suggesting in a preliminary way that side shear 

development was essentially frictional. This speculation is reinforced 
by the observation that excess pore pressures were very small through- 
out the test. 

A similar set of curves was developed for the average of the group 
piles. These curves are displayed in Figs. 3.15 and 3.16. (Only the 
average variation of residual stress among group piles was considered 
due to scatter in the data. See Chapter 1 and Appendix E.) State- 
ments concerning residual stress effects relevant to the references piles 
are also generally valid for the group piles. The negative residual 
shear stresses in the upper portions of the piles tended to be lower in 
the group piles than in the reference piles. In the upper 15 ft. (4.6 
m) the residual side shear stresses were positive, presumably due to 
shearing resistance produced by the weight of the cap. In the bottom 
several feet of the group piles the residual side shear was higher than 
at the corresponding depths in the reference piles. When this residual 
stress is added to the apparent stresses, the true maximum shear 
stresses and shear stresses at pile failure are seen to be higher for the 
group piles than for the reference piles below 35 ft. (10.7 m). This 
phenomenon is believed to be due to the increasingly granular nature of 
the sandy clay in Zone D below that depth. 

The no-load zeros could not be maintained beyond Test 1 due to 
drift in the data collection system, thought to be due primarily to 
intrusion of minute amounts of water into the lead wire. See Appendix 
E. In order to attempt to estimate the variation of true unit side 
resistance by soil layer between the first and last compression tests, 
the true load distributions, including the effects of residual stresses, 
at failure were calculated for the last compressive loading by using a 
procedure suggested by Hunter and Davisson (Ref. 32, Interim Report ). 
In this procedure an apparent load-depth curve at failure is corrected 
by subtracting from a given measured apparent load value at failure the 
apparent load value at that level observed upon unloading and adding 
to the result the absolute value of the apparent load remaining in the 
pile at the same level after conducting an uplift load test to failure and 

148 



-30 - 



20 

i— 



RESIDUAL LOAD CIO 
10 10 20 



30 40 



LOWER 
BOUND 



1 



NOTE: BOUNDS 
NOT DEVELOPED 
BY ANY SPECIFIC 
PILE 





UPPER 
BOUND 



AVERAGE 
AFTER 
> TEST I 



FIGURE 3.15. AVERAGE RESIDUAL LOADS IN GROUP PILES; TEST 1 
(1 ft = 0.305 m; 1 k = 4.45 kN) 



149 



-500 



f CPSFD 
1000 1500 



2000 2500 




AT FAILURE: 
CONSIDERING 
RESIDUAL STRESS 
EFFECTS CAV6. 
PRE-TEST RESIDUAL 
SIDE SHEAR 
STRESS ON 
GROUP PILES D 



j^-BEFORE TEST: 

\ cavg. residual 
Aside shear stress 
\on group pilesd 

+ 



AT MAX SIDE SHEAR: 
WITHOUT CONSIDERING 
RESIDUAL STRESS EFFECTS 



AT FAILURE: WITHOUT 
CONSIDERING RESIDUAL 
STRESS EFFECTS 



AT MAX SIDE SHEAR: 
CONSIDERING 
RESIDUAL 
STRESS 
* EFFECTS 



43 L * 



DOES NOT OCCUR AT SAME LOAD IN EVERY 
PILE OR AT EVERY DEPTH 

FIGURE 3.16. AVERAGE F-D RELATIONSHIPS FOR GROUP PILES; TEST 1 
(1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



150 



unloading. Loads are zeroed before the uplift test. The resulting 
adjusted load distribution diagram should approximate the true load 
distribution at failure. A similar type of correction procedure was used 
to adjust the apparent load distribution diagrams at failure in uplift to 
produce a true load distribution diagram for uplift behavior. 

The average apparent and adjusted load distribution diagrams for 
the two reference piles are given in Fig. 3.17. The compression tests 
represented are the last tests, conducted in conjunction with the final 
9-pile test. Similar average diagrams for Group Piles 2, 4, 5, and 9 
(those tested in uplift) are given in Fig. 3.18. In the case of the 
group piles the compression load-depth diagram for each individual pile 
was taken as that corresponding to the last compression test conducted 
(e.g., last 9-pile test for Pile 4; 4-pile test for Pile 5). 

Differentiation of the load distribution diagrams in Figs. 3.17 and 
3.18 yields the unit side shear distribution at total pile failure for the 
last compressive loading of each pile and for uplift loading that followed 
the several cycles of compression loading. The results of this differ- 
entiation, expressed as average values of unit side shear at failure, are 
tabulated in Table 3.3. For purposes of comparison, values obtained 
from direct measurement on Piles 2, 4, 5, and 9 for the first compressive 
loading (first 9-pile test) are also shown. 

Conclusions that can be drawn from this table are that (1) no 
significant changes occurred in the reference piles in true (adjusted) 
unit side resistance in any of the four soil zones between the first and 
last (third) compressive loadings; (2) significantly reduced true side 
load transfer occurred in the reference piles in the more granular soils 
below a depth of 26 ft. (7.9 m) in uplift as compared to the last com- 
pressive loading and (3) reduction in true unit side shear stresses at 
failure occurred in the group piles between the first and last compres- 
sive loading in Zones B and C, in the depth range of 11 to 31 ft. (3.4 
to 9.5 m). Unexpectedly, these losses in Zones B and C were recovered 
when the group piles were subjected to uplift tests. This phenomenon 
may be due to fabric reorientation in the highly plastic Beaumont clay 
that results from a reversal of the direction of applied shearing strain. 
The group piles developed slightly higher true side shear stresses at 
failure than did the reference piles in the more granular deeper soil 
zones . 

It should be emphasized that the side shears actually available to 
the piles to resist applied load are those tabulated in Table 3.2 and not 
the true values in Table 3.3. Table 3.3 does, however, provide some 
insights into the way in which the true shear strength of the soil 
varies with loading cycle and direction of loading. 

Unit Load Transfer Curves 

Graphs of average developed side shear stress versus downward 
pile deflection (f-z curves) are presented for the four principal soil 
zones for the reference and group piles, respectively, in Figs. 3.19-3.22. 

151 



-100 -50 



INDICATED LOAD (K) 
50 100 



200 




'H(FT) AT 
INTERVALS 



LEGEND 
• TEST 3, MAX LOAD 
oTEST 3, UNLOADED 
a UPLIFT, MAX LOAD 
a UPLIFT, UNLOADED 

■ADJUSTED TEST 3, 
MAX LOAD 

° ADJUSTED UPLIFT, 
MAX LOAD 



FIGURE 3.17. AVERAGE APPARENT AND ADJUSTED LOAD DISTRIBUTION DIAGRAMS 
AT FAILURE FOR REFERENCE PILES (1 ft = 0.305 m; 1 k = 4.45 kN) 

152 



INDICATED LOAD ( K ) 
50 100 




DEPTH (FT) AT 
5 FT INTERVALS 



LEGEND 

MAX 
• COMPRESSION, LOAD 

©COMPRESSION, lo^ed 

A UPLIFT, MAX LOAD 

AUPLIFT, UNLOADED 
■ ADJUSTED COMPRES- 
SION, MAX LOAD 
D ADJUSTED UPLIFT, 
MAX LOAD 



FIGURE 3.18. AVERAGE APPARENT AND ADJUSTED LOAD DISTRIBUTION DIAGRAMS AT 
FAILURE FOR GROUP PILES 2, 4, 5, AND 9 (1 ft = 0.305 m; 1 k = 4.45 kN) 

153 



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(ft) 


o-ii 

11-26 
26-31 

31-42 



<D 
IO 

0- 

cvj" 

CO 

h- 
co 

LU 



3 



Q 
LU 

= = = §■« 

=> hr 

CO t 



CO 
LU 
CO 
CO 
LU 



LU 

_J 

Q_ 

(L 

3 
O 



< 

3 

9 
co 

LU 



or or 
cr> 



Li. 

o 



or 
o 



o 

LU 



co 

-J 
< 

3 

9 
co 

LU 

or 

or 
o 

Ll 



LU CO LU 



cr> 
< 
or 

LU 



3 

a 
< 



h- 
co 

3 

Q 
< 



o jo o 



LU^T 
>CM 

co9 

LUZ 

or< 

2LU 

oz 

LU 
HOT 
COLU 
<Ll 
_l LU 

or 

d = 

z — 

I- co 
c^y 

LUd 

hi 

CO DC 

<o 



t cocoor 55" 



CDLU 
Z-J 

QCD 



LUf° 

>b 
coz 

Qjl - 

or co 

^h- 



O 



O 



Q .=: " 



EO 



LU 

I- 
O 



or 

£*r 

cog 

bs . 

=><co 

CO^H 
LULU CO 

or or lu 
Qlu 1- 

UJXh- 

<co^ 

LU 2 

cd or o 

10 CO Li. 

y<Lu 

— 30 

0-93 
or cop 

qLU LU 
l-O LU 

coLUor 
luot < 

1-3 

n coco 

O5 3 

or*^co 

CDQ LU 

qqlu or 
C03Q 

dz CO 

0_— 3 

Ch-§ 
3CO^" 
OLULu 

LLhO 



154 



z/z 



REFERENCE PILES 

3.0 AND 8.5 FT LEVELS 

TEST I (LAYER A) 



f(psi) 




(I) 



f/f 



max 



-5 



0.1 0.2 

(2) Z(in.) 



(1) NEGLECTING RESIDUAL STRESSES 

(2) INCLUDING RESIDUAL STRESSES 



f(psi) 




REFERENCE PILES 

13.5 FT, 18.5 FT, AND 23.5 FT 

LEVELS 

TEST I (LAYER B) 



-5 



(I) 



max 



0.1 0.2 

Z(in.) 

(1) NEGLECTING RESIDUAL STRESSES 

(2) INCLUDING RESIDUAL STRESSES 



FIGURE 3.19. F-Z CURVES; SOIL ZONES A AND B; REFERENCE PILES; TEST 1 
( 1 f t = 0.305 m; 1 in = 25.4 mm; 1 psi = 6.89 kN/m ) 



155 



z/z 



10 



f(psi) 



f(psi) 



1 


2 3 












REFERENCE PILES 
28.5 FT LEVEL 
TEST 1 ( LAYER C) 


1 1 


1 1 


-m 


►1 


(1) 


— I 


(1) 
0-5 f/f mQX 


¥ 


0.1 

Z(in 


0.2 
.) 



(1) NEGLECTING RESIDUAL STRESSES 

(2) INCLUDING RESIDUAL STRESSES 



z/z, 

I 2 3 



(^O^-O 



10 



(2) 




(I) 



REFERENCE PILES 

33.5 FT, 37.5 FT, AND 40.5 FT 

LEVELS 

TEST I (LAYER D) 



(I) 

1 Q5 f/f max 



FIGURE 3.20, 



0.1 0.2 

2 (in.) 

(1) NEGLECTING RESIDUAL STRESSES 

(2) INCLUDING RESIDUAL STRESSES 

F-Z CURVES; SOIL ZONES C AND D; REFERENCE PILES; TEST 1 
( 1 ft = 0.305 m; 1 in = 25.4 mm; 1 psi = 6.89 kN/m ) 



156 



f(psi) 





z/z c 

1.5 2 




ALL GROUP PILES 

3.0 AND 8.5 FT LEVELS 

TEST 1 (LAYER A) 


0.5 1 








i 

i 
i 

i 


I 1 


(21 


k^tfr: 


2 (l) 

d f/f 

, "'max 




1 




I 


0.1 


0.2 
Z(in.) 


0.3 0.4 



f(psi) 



(1) NEGLECTING RESIDUAL STRESSES 

(2) CONSIDERING RESIDUAL STRESSES (POSITIVE 
VALUES AT THIS LEVEL DUE TO CAP WEIGHT) 

0.5 Z/Z c I 




ALL GROUP PILES 
13.5 FT, 18.5 FT AND 
23.5 FT LEVELS 
TEST I (LAYER B) 



(I) 



max 



(1) NEGLECTING RESIDUAL STRESSES 

(2) CONSIDERING RESIDUAL STRESSES 



FIGURE 3.21. 



F-Z CURVES; SOIL ZONES A AND B; GROUP PILES; TEST 1 
( 1 f t = 0.305 m; 1 in = 25.4 mm; 1 psi = 6.89 kN/m ) 



157 



10 



z/z c 

0.5 



5 - 



f(psi) 



1 

(1) 

h 


n 


(2). 



ALL GROUP PILES 
28.5 FT LEVEL 
TEST I ( LAYER C) 



05 f/f 



(I) 



max 



0.1 



-5 



0.2 
Z(in.) 



0.3 



15 



(1) NEGLECTING RESIDUAL STRESSES 

(2) CONSIDERING RESIDUAL STRESSES 

z/z c 

0.5 I 1.5 2 



10- 



f(psi) 



5 - 



' 1 

" if 

il 


i 1 
— -o 


(1) 


i 


1 



ALL 


GROUP PILES 






33.5 


FT, 


37.5 FT, AND 


40.5 


FT 


LEVELS 








TEST 


1 


(LAYER D) 







(I) 

H0.5 f/f max 



0.0 



0.1 



0.2 



FIGURE 3.22 



Z(in.) 

(1) NEGLECTING RESIDUAL STRESSES 

(2) CONSIDERING RESIDUAL STRESSES 

F-Z CURVES; SOIL ZONES C AND D; GROUP PILES; TEST 1 
( 1 f t = 0.305 m; 1 in = 25.4 mm; 1 psi = 6.89 kN/m ) 



158 



These graphs refer to the first test set on the reference piles and 
9-pile group, and in general each represents the average of f-z relation- 
ships developed at two or three stations between strain gage levels. 
Similar relationships for the remaining tests may be found in Appendix 
D. The f-values were obtained by differentiating fitted f-d curves at a 
given depth (e.g., 3.0 ft. (0.92 m)) for each value of applied load. 
Corresponding z-values were obtained by subtracting the elastic 
compres sion between the level of the settlement gages and the depth in 
question from the average settlement gage reading for the pile under 
consideration. The elastic compression was computed from the area 
under the partial measured load distribution versus depth curve divided 
by elastic stiffness. 

These curves are instructive in visualizing the buildup of unit side 
shear with pile displacement in the various zones of soil, defined in 
Fig. 1.1. The curves for the reference piles are also necessary inputs 
into the pile group behavior model described in Chapter 4. The 
various f-z curves have been plotted considering both pretest zeros 
(apparent curves, which neglect residual stresses) and predrive zeros 
(true curves, which include the effects of residual stresses) and have 
also been plotted in normalized form, in which z is the displacement 
corresponding to peak load transfer. Several observations may be made 
concerning these curves: (1) the critical displacements in the reference 
piles are much smaller than had been anticipated, ranging from about 
0.05 in. (1.3 mm) in Zones B and C to about 0.02 in. (0.5 mm) in Zone 
D, near the bottoms of the piles; (2) the average critical displacements 
for the piles in the 9-pile group ranged from about 0.2 in. (5.1 mm) in 
Zones B and C to about 0.1 in. (2.5 mm) in Zones A and D; (3) the 
differences between the curves which include residual stresses and 
those which neglect residual stresses could be classified as relatively 
minor . 

Tip load versus tip movement curves (Q-z curves) for Test 1, 
derived from measured tip load (Q) and values of tip movement (z) 
computed by the method employed for computing z-values for the f-z 
curves, are shown in Fig. 3.23. Corrections to the z-values were made 
for the reference pile tests because of the very small movements 
associated with small tip loads coupled with small errors in measuring 
pile settlement (and thereby z at the tip) produced by surface tem- 
perature variations and computed pile flexibility. 



Load Transfer Correlations 

Correlations of apparent measured maximum unit side resistance 
values to soil shear strength obtained by several methods are presented 
in Table 3.4 for Test 1. Similar correlations for the other tests are 
presented in Appendix F. The correlations, which do not include 
residual stress effects, are ratios of maximum observed side shear 
stress to: 



159 







• NEGLECTNG RESIDUAL TP LOAD 
O CONSIDERING RESIDUAL TIP LOAD 



OFFSET 

TO ERRORS 

IN SETTLEMENT 

ASSOCIATED 

WITH TEMPERATURE 

CHANGES 

30 r 



0.2 
Z(IN.) 



0.4 



UJ 

.j 

Q. 

q: 

IxJ 
Q. 

o 




COMPLETE FAILURE 

NOT ACHIEVED IN ALL PILES 

• NEGLECTING RESIDUAL TIP LOAD 
O CONSIDERING RESIDUAL TIP UOAD 



0.2 0.3 

Z(IN.) 



0.4 



FIGURE 3.23. Q-Z CURVES FOR TEST 1; REFERENCE PILES (ABOVE); GROUP 
PILES (BELOW) (1 k = 4.45 kN; 1 in = 25.4 mm) 



160 



TABLE 3.4. SIDE RESISTANCE CORRELATION FACTORS, 

9-PILE TEST 1 (1 FT = 0.305 m) 



FACTOR 


AVERAGE OF 


AVERAGE OF 


! (REFERENCE SOIL TEST) 


REFERENCE PILES 


GROUP PILES 


a (UU TRIAX.) 






STRATUM A 


0.08 


0.07 


B 


0.65 


0.58 


C 


0.67 


0.68 


D 


0.54 


0.45 


OVERALL 


0.48 


0.44 


a RC (REMOLDED UU TRIAX.) 


0.47 


0.43 


a (CONE SLEEVE) 






STRATUM A 


0.16 


0.16 


B 


0.96 


0.86 


C 


1.21 


1.23 


D 


1.48 


1.37 


OVERALL 


0.95 


0.88 


a Lp (LIMIT PRESSURE) 


0.0252 


0.0234 


a (CU TRIAX. W/ MEASURED 






PRETEST EFF. STRESS) 






DEPTH 9' 


1.89 


0.25 


19 ' 


1.61 


0.57 


34' 


0.33 


0.27 


41' 


0.47 


0.58 


a (CU TRIAX. W/ MEASURED 






EFF. STRESS AT FAILURE) 






DEPTH 9' 


1.66 


0.30 


19' 


2.37 


0.54 


34' 


0.61 


0.33 


41' 


0.60 


1.23 


A (UU TRIAX.) 


0.174 


0.162 


■^mov (measured) 






max v ' 






W GESM > 






DEPTH 10' 


0.49 


0.32 


20' 


0.87 


0.76 


35 • 


0.64 


0.68 


40' 


0.92 


0.92 



NOTE: Residual Stress Effects and Cap Weight Not Considered in Computations 

161 



1. Shear strengths from UU triaxial compression tests on un- 
disturbed samples (Fig. 1.2), resulting in the a factor. 

2. Shear strengths from UU triaxial compression tests on 
remolded samples ( Interim Report ) , resulting in the a R ~ 
factor. KU 

3. Shear strengths as indicated by sleeve friction on the static 
cone penetrometer. 

4. The average limit pressure to a depth of 43 ft. (13.1 m) from 
the self-boring pressuremeter, yielding the a T p factor. 

5. The peak shear strengths inferred from the CU triaxial 
compression tests, in which the confining pressure equals the 
measured pretest lateral effective stress on the pile face. CU 
Mohr- Coulomb envelopes from Appendix C were used to assess 
these strengths. 



6. The peak shear strengths inferred from the CU triaxial com- 
pression tests, in which the confining pressure equals the 
lateral effective stress measured at failure. Composite 
effective stress Mohr-Coulomb envelopes for each zone from 
Appendix C were used to obtain these shear strengths. The 
ratio obtained from this correlation is denoted the a factor. 

7. The average UU triaxial shear strength to a depth of 43 ft. 
(13.1 m) plus twice the average vertical quasi-effective stress 
between the surface and that depth, denoted the A factor. 

The ratio of maximum measured side shear stress to maximum side 
shear stress computed by a version of the General Effective Stress 
Method (GESM) (Ref. 5 of Appendix A) is also given in Table 3.4. 

Correlations 1, 3, 5, 6, and the GESM correlation were made for 
the reference tests and the group test for each of the four principal 
soil zones or at four distinct depths. For Correlations 5 and 6, they 
were made using the measured effective stresses on the piles at the 
indicated depths and the shear strength parameters shown in Table 3.5, 
which were interpreted from the data in Appendix C (although they are 
not exactly equal to the cohesion and internal friction values tabulated 
in that Appendix). The remaining three correlations were average 
correlations over the entire embedded lengths of the reference and 
group piles. 

The following comments are offered concerning the correlations: 

1. The best overall correlation was with the cone sleeve, although 
it overpredicted load transfer in Zone (Stratum) A near the 
top of the pile and underpredicted load transfer in Zones C 
and D, near the bottoms of the piles. 

162 



TABLE 3.5. INTERPRETED PEAK COHESION (c) AND ANGLE OF INTERNAL 
FRICTION (<j>) VALUES USED FOR LOAD TRANSFER CORRELATIONS 
(1 ft = 0.305 m; 1 psf = 47.9 N/m 2 ) 



Depth 
(ft) 


Cohesion (psf) 


Angle of Internal Friction 
(degrees) 


c(CU) 


c(CU) 


<J>(CU) 


<HCU) 


9 


440 


500 


25 


22 


19 


300 





19 


23 


34 


2,500 


200 


20 


28 


41 


2,500 





20 


27 



163 



2. The a (UU triaxial) correlation was near the median value of 
0.45 to 0.50 currently used in stiff to very stiff clay by 
designers. Correlation to the remolded undrained shear 
strength was almost identical to that for undisturbed un- 
drained strength due to the characteristically underrepre- 
sented undisturbed values produced by sample disturbance 
and preferential failure along fissure planes. 

3. The K correlation was below the nominal value of 0.22 
generally recommended for a 43 ft. (13.1 m) penetration. 

4. The correlations with measured preload effective stresses and 
effective stresses at failure were not especially good. This 
may be due largly to the fact that the soil strength at failure 
may better represented by a residual strength from a direct 
or simple shear test than by peak strengths obtained from 
triaxial compression tests. 

5. The GESM predicted values of load transfer that were 
generally too large, although the depthwise accuracy achieved 
was generally equivalent to that achieved with the cone sleeve. 

An additional rational correlation, not addressed in Table 3.4 but 
referred to above, is the ratio of measured maximum unit side load 
transfer to shear strength computed by the product of the measured 
effective stress times the tangent of the residual angle of internal 
friction. This was not done because the residual friction angles 
reported in Appendix C are unrepresentative of the soils at the site, 
possibly because complete residual conditions had not been achieved in 
the laboratory tests. Representative values are believed to be approxi- 
mately as follows: 



<j> = 17 deg. at 9 ft. 

= 13 deg. at 19 ft. 

= 22 deg. at 34 ft. 

= 25 deg. at 41 ft. 



(2.75 m) depth 
(5.80 m) depth 
(10.37 m) depth 
(12.51 m) depth, 



If the tangents of the above angles are multiplied times the average 
measured effective stress at failure at each depth in all five piles 
instrumented for lateral effective stress and the result divided into the 
average measured apparent unit side resistances in those five piles for 
the first load test, the following dimensionless correlation factors result: 



0.90 at 9 ft. 

1.27 at 19 ft. 

0.49 at 34 ft. 

0.82 at 41 ft. 



(2.75 m) depth 
(5.80 m) depth 
(10.37 m) depth 
(12.51 m) depth 



The above factors do not consider effects of residual stress, which 
if included, should cause the factors to tend more toward unity. The 



164 



generally good correlations indicate that assessment of pile capacity in 
overconsolidated clay based on estimated lateral effective stress and 
measured residual shear strength could conceivably be reliable. 

Other correlations of possible interest to designers could be 
developed from the available data, including correlations with 
unconfined compressive strength, pocket penetrometer strength, and 
standard penetration test results. Such correlations, however, have 
not been included in this report. The unconfined compression and 
normalized parameter strengths should yield a factors in the same order 
as those developed using UU triaxial test results. 

Variability of Load Transfer 

The load transfer correlation factors shown on Table 3.4 and on 
the corresponding tables in Appendix F are average values for the 
reference or group piles for the soil zones (strata) indicated. The 
variability of the a factors based on UU triaxial and static cone test 
results from pile to pile and layer to layer can be assessed by 
multiplying the 6 factors from Table 3.6 times the average a factor (for 
reference or group pile, as appropriate) given for the stratum of 
concern in Table 3.4. For example, the a factor (UU triaxial) for Pile 
3, a group pile, in Stratum B was 0.96 (Table 3.6) times 0.58 (Table 
3.4), or 0.56. Analysis of Table 3.6 reveals that the a factors (and 
therefore unit side load transfer) was relatively uniform among the 
group piles, except in Stratum A. 

Load Transfer Correlation Factors at Pile Tips 

Correlations of maximum average developed tip load in the 
reference and group piles to UU triaxial shear strength and to peak 
static cone tip resistance in the soil immediately below the pile tips are 
given in Table 3.7. 

End bearing correlations, which yield end bearing capacity factors 
with respect to the tests mentioned, were made with and without 
consideration of residual tip loads. Measured residual tip loads were 
used for Test 1; computed loads based on the procedure described 
earlier for assessing residual side shear were used for Test 3 for the 
reference piles and for the last compressive loading on the group piles. 
The tip loads used to develop these factors are actual loads measured 
one ft. (0.305 m) above the tips. It has been tacitly assumed that no 
side resistance existed in the bottom one ft. (0.305 m) (approximately 
one diameter) of the piles. If load were in fact transferred in this 
zone, the factors given in Table 3.7 would be too high. 

Correlations with UU triaxial test results yielded unrealistically 
high factors, probably due to two effects: (1) sample disturbence and 
(2) frictional behavior along with full or partial drainage in the soil 
beneath the pile tips, due to the high sand content of that soil, that 



165 



TABLE 3.6. FACTORS (5) FOR COMPUTING a CORRELATIONS FOR INDIVIDUAL 
PILES AND LAYERS, 9-PILE TEST 1 (1 ft = 0.305 m) 



PILE 


STRATUM 

A 
(0 - 10') 


STRATUM 

B 

(10' - 26') 


STRATUM 

C 

(26' - 30') 


STRATUM 

D 

(30' - 43') 


1 (REF) 


1.07 


1.10 


1.11 


1.16 


2 


0.82 


0.99 


1.07 


1.04 


3 


1.59 


0.96 


0.90 


0.91 


4 


1.70 


1.06 


0.93 


0.90 


5 


0.73 


0.90 


0.87 


1.24 


6 


0.29 


0.90 


1.12 


1.27 


7 


0.82 


0.93 


1.00 


1.02 


8 


1.22 


0.94 


0.78 


0.79 


9 


1.12 


1.12 


1.14 


0.83 


10 


0.71 


1.20 


1.20 


0.99 


11 (REF) 


0.93 


0.90 


0.89 


0.84 



NOTE: a for a particular pile and layer = <S from above table times 
a from Table of Average Correlation Factors. This does not 
apply to a (LIMIT PRESSURE) or X. 



6 = 



MEASURED LOAD TRANSFER FOR PILE /STRATUM 

AVG. LOAD TRANS. FOR STRATUM FOR REF. OR GROUP PILES 



166 



TABLE 3.7. AVERAGE END BEARING CAPACITY FACTORS FOR 
REFERENCE AND GROUP PILES 



Test, Piles 


UU Triaxial/ 1 ' 
Neglecting 
Residual 
Tip Load 


Cone Tip/ ^ 
Neglecting 
Residual 
Tip Load 


UU Triaxial/ 1 ' 
Including 
Residual 
Tip Load 


Cone Tip, *■ J 
Including 
Residual 
Tip Load 


1, Reference 
Piles 

1, Group ^ 
Piles 

2, Reference 
Piles 

2 , Group 

Piles 

(4) 
3, Reference 

Piles 

Last Loading 
(4) (5) 
Group Piles 


31.8 
31.8 
24.3 
32.2 
21.6 
23.1 


0.65 
0.65 
0.50 
0.66 
0.42 
0.45 


37.5 
35.0 

47.2 
56.4 


0.77 
0.72 

0.93 
1.11 



(1) Peak unit end bearing stress divided by undrained cohesion at 
43-45 ft. (13.1-13.7 m) from UU triaxial test. 

(2) Peak unit end bearing stress divided by average peak tip bear- 
ing stress at 43-45 ft. (13.1-13.7 m) registered by static cone. 

(3) Includes only those piles where tip failure occurred. 

(4) Residual load values obtained by adjustment procedure involving 
uplift tests. 

(5) Includes only group piles subjected to uplift tests (2, 4, 5, 9). 
Last Loading taken as 4-pile Subgroup test for Piles 5 and 9 and 
third 9-pile Test for Piles 2 and 4. 



167 



could not be properly represented by the undrained triaxial tests. The 
cone tip correlations were much better. A correlation factor of about 
0.65 (measured on pile / cone tip reading) appears appropriate for 
apparent capacity upon first loading for both the group and reference 
piles. The correlation factor approached 1.0 after several loadings 
(total gross tip movement of 20-40 percent of pile diameter) when 
residual load effects were considered. 



168 



Chapter 4. Reanalysis of Performance 
Using Hybrid Model 



Introduction 



The hybrid method of pile group analysis has been described in 
some detail in the Interim Report and in Appendix A of this report. 
Chapter 4 of the Interim Report contained a prediction of the response 
of the test group, assuming certain ideal conditions and preliminary 
estimates of pile penetrations and soil properties using Program GP3B, 
an early algorithmic version of the hybrid model. An improved version 
of the hybrid model, Program PILGP1 (described in Appendices A and 
B), was used to model the performance of the test group upon 
completion of the field tests. This analysis, which is described in this 
chapter, used results of the site investigation performed for the study 
and as-built pile penetrations. Two solutions were obtained: (1) Using 
unit load transfer curves measured from the reference piles and 
complete as-built geometry (including inadvertant batters and true pile 
spacing) as inputs (hereafter called the "reanalyzed" solution) and (2) 
Using unit load transfer curves developed from criteria and ideal 
(as-planned) spacing and zero batters as inputs, hereafter called the 
"criterion" solution. The criterion solution represents a solution that 
could be made by a designer with minimal data . For the criterion 
solution, the following criteria (refer to Appendix A) were applied: 



f-z Curves: 



where 



f = f [2(z/z ) 

max c 

f = ac 
max u 



0.5 



z/z c ] 



in which 



%! [0.67 + An MUzY!] 
G r 



f = unit side resistance 

f „ = maximum unit side resistance 

max 

z = pile deflection 

z = pile deflection corresponding to f 



u 



r 



max 

= undrained shear strength from triaxial 
compression tests 

= proportion factor (=0.5 for most of soil profile) 

= pile radius 



169 






Q-z curve: 


Q 


where 


Q 




Q 




z 




z c 


Tests Modeled 





v = Poisson's ratio of soil (=0.5) 

G = Shear modulus of soil 

£ = Pile penetration (=43 ft or 13.1 m) 

p = G at middepth of pile/G at pile tip. 

For this analysis G was taken as that value given by the self-boring 
pressuremeter at the various depths at which f-z curves were input. G 
was observed to be about 450 times the UU triaxial shear strength of 
the soil. 

« = Q max (z/V°- 33 

= end bearing force 

= bearing capacity 

= 9 x tip area x UU triaxial shear strength 
of soil at elevation of pile tips 

= tip deflection 

= 0.03 x tip diameter. 



For purposes of examining the capabilities of the model to predict 
pile group performance, 9-pile Test 1, the 5-pile Subgroup Test, and 
the 4-pile Subgroup Test were modeled. The former test was modeled 
with the reanalyzed and the criterion solutions, while the latter tests 
were modeled only with the reanalyzed solution. 

Geometric Inputs 

A coordinate system was established with its origin (O) at the 
geometric center of the bottom of the pile cap, as shown in Fig. 4.1. 
Coordinates in the X-Z plane were then established for the pile heads, 
assumed to be at the base of the rigid cap. These coordinates for the 
reanalyzed solution, also shown in Fig. 4.1, were taken to be identical 
to the coordinates at the actual pile tops, reported in Chapter 1. For 
the reanalyzed solution, the direction angles for the piles, as projected 
onto the X-Z plane, were taken as shown in Fig. 4.2 and are based on 
measurements reported in Chapter 1. The true batter slopes, or base 
offsets divided by pile lengths, also input into PILGP1, are tabulated in 
Table 4.1, which additionally gives tabular values of pile head 
coordinates and direction angles. For the criterion solution, the center 
of the pile cap coincided with Pile 2, and the unbattered piles were 
taken to be on a 32.25 in. (819 mm) grid. 



170 



\ 



1 




N ,0 t 


1 



-30.9' 



-J 



30.4" 



-33.0 



1.8' 



m, 



^j 



30.0" 



■1.8' 



CAP BOUNDARY 



31.8" 



•2.0' 



-32.0" 



33.0" 



0.5"-H 



29.8' 



33.0" 



+ -i- 



-33.5 



TV 



FT 

2.0 



31.0" 



J- + 6 



~l 



FIGURE 4.1. PILE HEAD COORDINATES FOR PILGP1 ANALYSIS (1 in = 25.4 mm) 



171 



10 



329* 




316° 




8 




220° 

3 



340 





345' 




^ 



350 ( 




FIGURE 4.2. DIRECTION ANGLES 
172 



TABLE 4.1. PILE GEOMETRY FOR PILGP1 REANALYSIS (1 in. = 25.4 mm) 



Pile 


Direction angle 
a (deg.) 


True batter 
slope 


Pile head coordinates 


(in.) 


X 


Y 


z 


2 


340 


0.014 


- 0.8 


0.0 


- 1.8 




3 


220 


0.009 


-32.0 


0.0 


- 1.8 




! 4 


345 


0.013 


-33.0 


0.0 


30.0 




5 


78 


0.008 


- 0.5 


0.0 


29.8 




6 


350 


0.014 


33.0 


0.0 


31.0 




7 


145 


0.020 


32.0 


0.0 


- 2.0 




8 


316 


0.033 


31.8 


0.0 


-33.5 




9 


329 


0.022 


- 2.0 


0.0 


-33.0 




10 


191 


0.007 


-30.9 


0.0 


-30.4 





173 



Loadings 

The loading inputs for the reanalyzed solution were developed from 
the loads measured at each of the four jack locations. Variation of load 
from jack to jack and slight positioning anomalies for the jacks were 
taken into account by also inputting moments about the X and Z axes 
equal to the sums of the products of the jack loads and the moment 
arms depicted in Fig. 4.3. Multiple loadings, not associated with 
measured loads, were also run in order to develop the entire load- 
settlement curve. Cap weight was not included in the applied loads. 
Similar considerations were made for the criterion solution, except that 
no load eccentricity was used. 

Structural Properties 

Each pile was assigned a cross-sectional area of 11.91 square in. 
(7884 square mm) and a Young's modulus of 30,000,000 psi (206,000 
mN/m 2 ). Moments of inertia were computed internally within the 
program from the inside and outside diameters of the piles (10.02 and 
10.75 in., respectively) (255 and 273 mm, respectively). Each pile was 
specified to protrude 3 ft. (0.915 m) above the ground surface. Forty- 
six discrete elements were used to represent the piles. 

Soil Inputs 

For unit side shear, four'f-z curves were assumed to represent 
soil response. These curves are shown in Fig. 4.4. Each specific 
curve for the reanalyzed solution was developed by averaging the 
experimental f-z curves (obtained at 5-ft. (1.53 m) depth intervals 
beginning at a depth of 3 ft. (0.92 m)) within the depth intervals 
noted from Piles 1 and 11. The Q-z curve for the reanalyzed solutions, 
Fig. 4.5, is the average corrected curve for the reference piles. Soil 
inputs for the criterion solution are described on pp. 169-170. Exact 
inputs for the unit load transfer curves are tabulated in Tables 4.2 and 
4.3. 

Since the piles were nearly vertical, lateral unit soil resistance 
curves (p-y curves) were not input, except for one run to compare 
results obtained with and without inputting these curves. For this 
special run the non-cyclic stiff clay criteria described in Appendix A 
were employed to develop p-y curves. 

For purposes of making group effect calculations the Poisson's 
ratio of the soil was taken at 0.5 (incompressible soil) and the variation 
of Young's modules (E) with depth was taken in accordance with the 
relationship shown in Fig. 1.3 for the in-situ pressuremeter . Several 
other, constant values of E were also input in order to assess the 
sensitivity of the solution to the choice of E and to obtain the optimum 
value with respect to prediction of load-settlement and load transfer. 



174 



t 

N 



L 



t r 

-18.4" ., 6-9 « 



I*— 16.4' 
o 2 



164" 



3 o -,- 

1 i r-. nl< 



-16.7 



N -18.2' 



15.0" 



16.5" 



1 i:*i 

|-e--|6.6"- 



! [ 



15.5" l4 0" 

a 1 



16.8 



z 



O Position for subgroup tests 



+ Position for 9~pile test no. I 



1 



FIGURE 4.3. JACK COORDINATES (1 in = 25.4 mm) 



175 



5r 
f(psi) 





,«V 




CRITERION CURVE 

(0 - 9.99 feet) 

REF. PILE CURVE 



/ 

"♦-• 



i. 



0.1 

z(in.) 



0.2 




(2600-29.99 feet) 



0.1 0.2 

z(in.) 



IOr 




• — » m 
• — •--•^ 



(10.00-25.99 feet) 



0.1 0.2 

z(in.) 



I5 r 

10 
f(psi) 

5 







-• — •- 



(30.00-43.00 feet) 







0.1 0.2 

z(in.) 



FIGURE 4.4 F-Z CURVES FOR PILGP1 INPUT (1 psi - 6.89 kN/m ; 
1 ft. = 0.305 m; 1 in. = 25.4 mm) 



176 



REF. PILE CURVE 



0.1 



CRITERION CURVE 




02 

z (in.) 



03 04 



FIGURE 4.5. Q-Z CURVE FOR PILGP1 INPUT 
(1 k = 4.45 kN; 1 in. = 25.4 mm) 



177 



LO 

o 

tO 

o 
II 



+-> 

4-1 



LO 
CM 



o4 
w 
H 

h- 1 /— ■ \ 
Qicsl 

u g 
Q 2 

a oo 

w • 

>- 






a) 
z 

o 

E- 
O 



Oh 

I— I 
Cu 

04 

o 

CO 

w 
u 



4-1 



CM 



w 
< 



ii 



c/> 
ft 



/ 


















4-> 


















•H 4-> 


















U C-\<+H 


O 


CT> 


CM 


CM 


CM 


CM 


(Nl 


C\l 


O -H 


O 


00 


r^ 


r^ 


r^ 


[-- 


1^- 


t^ 


^ino 


















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o 


00 


CT> 


<y> 


CTl 


CT> 


CTi 


a> 


t+n w . 


















vO 


















rs ^T 


















C O 


















as o 


o 


o 


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o 


O 


LO 


LO 


LO 


<D /— X • 


o 


o 


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-sf 


to 


CM 


CM 


CM 


U -H to 


















v-^ (/) K> 


o 


I— 1 


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i—l 


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o 


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o 


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LO 


LO 


LO 


LO 


fn -H to 


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to 


00 


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o 


o 


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o 


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o 


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o 


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o 


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• 


o 


T-H 


CM 


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00 


o 


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o 


o 


o 


o 


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o 


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o 


■H 


















v 


o 


o 


o 


o 


O 


o 


o 


T-H 



•H 

P. 

in 
o 



o 
to 



p 

r— I 

3 

o 



o 



CM 

O 



cr 
to 

00 

o 
en 



a3 

^1 
< 



LO 





H 






0) 


+-> 




+-> 







<D 


e 




6 


rt 




aj 


•H 




•H 


Tl 




n3 


O 




0) 


•X3 




T3 


•H 




•H 


V) 




!/) 


+-> 




c 


P 




I—l 


o 


C/) 






UJ 






1 — 1 






H 






Oh 






W 






D- 






o 






04 






Oh 






LU 






J 







w 

H 
O 

2 



178 



TABLE 4.3. Q-z CURVES FOR PILGP1 SOLUTIONS 
(1 in. = 25.4 mm, 1 kip = 4.45 kN) 



; Z 


Q (rean.) 


Q (crit.) 


(in.) 


(Kips) 


(Kips) 


0.00 


0.00 


0.00 


0.02 


16.00 


14.40 


0.04 


24.80 


18.40 


0.08 


29.60 


22.80 


0.12 


33.00 


26.40 


0.22 


35.00 


32.00 


0.32 


35.70 


36.40 


1.00 


35.70 


36.40 



NOTE: Q (crit.) values were obtained by 
assigning cohesion values equal to 
4 times l r J triaxial values, since 
observation of profile reveals that 
UU triaxial strengths significantly 
underpredict in-situ strength in 
sandy clay near pile tips. 



179 



Results - Single Pile 

The single (reference) pile behavior produced by PILGP1 from the 
input described previously is represented in Figs. 4.6 and 4.7, which 
compare measured and predicted pile head load-settlement behavior and 
load distribution along the piles. In each case the "measured" values 
are averages of Piles 1 and 11. The "computed" values are for the 
reanalyzed solution and the "computed (cr)" values are for the criterion 
solution. The slight overprediction of capacity observed in Fig. 4.6 for 
the reanalyzed solution is due to the use of average f-z curves over 
the depth regions defined in Fig. 4.4 instead of using separate f-z 
curves for each of the 46 discrete, elements. Some further error was 
introduced by using f-z curves that were derived from fitted, rather 
than raw, load distribution diagrams. The same general comments apply 
to the comparisons of load distribution shown in Fig. 4.7, which was 
made only for the reanalyzed case. Theoretically, the measured and 
computed load distribution curves should be identical since the program 
is modeling the test from which the inputs were derived. 

Based on the comparisons shown in Figs. 4.6 and 4.7, the unit 
load transfer inputs were judged appropriate to model the group tests. 
Since the computed single pile capacity from the reanalyzed case 
exceeded the average measured failure load by an amount almost equal 
to the set-up that occurred between the first and final tests, no 
alterations in the f-z or Q-z curves were made for purposes of modeling 
the two subgroup tests. 

Results - Pile Groups 

9-Pile Test 1. Comparative plots of the load-settlement curves 
obtained for 9-pile Group Test 1 are presented in Fig. 4.8. The 
curves shown on that figure are the measured load-settlement curve, 
the reanalyzed load-settlement curve obtained by using the previously 
described f-z and Q-z curves and no p-y curves for the variation in 
soil modulus indicated by the pressuremeter, the reanalyzed load- 
settlement curve obtained by using a uniform soil modulus of 25 ksi (172 
mN/m 2 ) and no p-y curves, and a repeat of the 25 ksi (172 mN/m 2 ) 
run for the reanalyzed case including p-y curves developed according 
to the stiff clay criteria. The criterion case is also shown for E=25 ksi 
(172 mN/m 2 ). 

It is obvious that inputting an elastic modulus (E) variation 
equivalent to that measured in-situ with the self-boring pressuremeter 
(2.5 ksi (17.2 mN/m 2 ) at the surface, varying linearly to 11.5 ksi (79.2 
mN/m 2 ) at the depth of the pile tips) produced a softer load-settlement 
relationship than that which was measured. Use of modulus values from 
the UU triaxial or normalized strength triaxial tests would have resulted 
in even greater discrepancies. The best match occurred when E use 
taken as a depthwise uniform value of 25 ksi (172 mN/m 2 ), which is 
slightly greater than twice the in-situ modulus in the soil immediately 



180 




CM 

d 
CNI) 



fO 5t 

b d 

1N3IN3"LLL3S 



d 



4 
(0 

d 



181 



20 

i 



LOAD (K) 
40 60 




MEASURED (AVG. OF 
PILES I AND II ) 



- COMPUTED ■■ PILGPI 



43 L 



FIGURE 4.7. COMPUTED AND MEASURED MEAN DISTRIBUTION OF LOAD; REFERENCE 
PILES; TEST 1; PILGPI fl k = 4.45 kN; 1 ft = 0.505 m) 

182 




E-i 
CO 
W 
H 

W 

i— i 

Oh 

I 

CTi 



.- «sfr 



CO 



OS 
CJ 






h e 

E- \ 
tu Z 
co ^ 
i 

a a* 

< 00 

o • 

Q II 

E— th 
3 (/> 
0- Ph 

O rH 

u 

Q Z 

Q '* 

2 «* 
z> 

co n 
<C 

tu ^ 
2 



00 

UJ 

oi 



(Nl) 1N3KI31H3S 



183 



below the pile tips as indicated by the pressuremeter but less than 
either the average E along the depth of the piles or the E just below 
the tips of the piles indicated by the crosshole seismic tests. The 
appropriateness of such a high value for E can be explained in terms of 
the low strain values measured in the soil around the group, the 
reinforcing effects of the piles on the soil not accounted for explicitly 
in the hybrid model, and the influence of the relatively stiff er soils 
beneath the pile tips on the load-settlement behavior. With respect to 
the low measured strain values, even the self-boring-type pressure- 
meter used in the field study requires some lateral straining to seat the 
expanding membrane firmly against the sides of the borehole, so that 
the reported moduli were not obtained at strain amplitudes as low as 
those produced in the soil outside the immediate vicinity of the 
individual piles in the group. 

The best-fit E value corresponds to an average E/c of 1400 when c 
is based on average UU triaxial test results to a depth of 45 ft. (13.7 
m). This ratio is consistent with the general range of values observed 
in the Interim Report for the BRE and AREA tests, which were also 
conducted in overconsolidated clay. Since the measured settlements may 
have been too low by perhaps 10 to 20 percent at low load values due 
to small movements of the reference system relative to the group, it 
may be assumed that a more appropriate E/c for predicting true load- 
settlement behavior for the soil at this site would be about 1200. This 
compares with the range of 400 to 800 originally assumed in the first 
analysis reported in Chapter 4 of the Interim Report . 

The dashed curve in Fig. 4.8 is also for a uniform E of 25 ksi 
(172 mN/m 2 ), but p-y curves were included as inputs for the computa- 
tions employed to develop that curve. It can be seen that the inclusion 
of p-y curves in a case where the piles are slightly battered, as 
occurred here, influences the vertical load-settlement behavior. When 
p-y curves are not input the model must generate pile-head stiffness 
terms artificially whenever batter piles exist. In the case of the stiff 
soil at this site these stiffnesses were slightly too soft. 

The overestimation of capacity in the reanalyzed solution was the 
result of the overestimation of single pile capacity produced by the use 
of average f-z curves, discussed earlier. 

Computed and measured settlement ratios are shown in Fig. 4.9. 
Program PILGP1 does not output settlement ratios directly but they can 
easily be deduced from the non-interactive axial mode curve (load- 
settlement tabulation for an isolated pile) and the computed group 
settlement. The value of computed settlement ratio is somewhat 
dependent on the exact formulation used for the f-z and Q-z curves, as 
can be seen by comparing criteria and reanalyzed results. Better 
matches in both the settlement ratios and load-settlement plots could 
have undoubtedly been obtained by manipulation of the input para- 
meters. However, this was not the objective of this analysis. 



184 



tr 
o 

Q 

UJ 

3 
Q. 
2 
O 
O 



o o o 

UJ UJ UJ 

<r h- a: 

-3 ? => 

co a. co 

< 5 < 

ui o uj 

2o 2 



UJ 

o 

UI 

_l 



CO 
UJ 



Q. 

3 

o 
tr 
e> 

UJ 

-i 



& 



ui 
a. 

O 

a: 
o 

s 

CO 
UI 

-J 

GL 

I 
if) 



° ° A 

LU UI UJ 
hCl- 

3 3 3 

Q_ CO CL 

2 < 2 

O UJ o 

O 2 O 



f3 

Q. 

3 
O 

a: 
o 
m 

3 
CO 

UI 



a. 
i 



t/ 



-i o 



o 


I— < 


a> 


E- 




co 




UJ 




E- 




UJ 


o 




CO 


a, 




a> 


o Q 


o 


fc- S 


1— H 

F-c 


o 


< 


-i 


OS 


^ LU 




8§ 


UJ 


-J 


E- 


8 J 


E- 

UJ 

Q 


3 


UJ 


O 


E- 

IT3 


a: 


O 


5t U. 




O 


Q 


H 




z 


a 


O UJ 


UJ 
OS 


fO o 




cr 


< 


UJ 


UJ 


Q_ 


o 




CM 


CTl 




Tf 




UJ 




as 


o 


3 
C5 



IX 



ro 



C\J 

0I1VU lN3l^l3nil3S 



185 



Computed distributions of loads to the pile heads are tabulated for 
the case of E = 25 ksi (172 mN/m 2 ) and no p-y input in Table 4.4, 
which also gives the measured values. In general, the correspondence 
is good, although some differences exist at a load of 1274.7 k (5.67 
mN), which was the measured failure load. The computed loads for the 
reanalyzed case are not symmetric because the program has accounted 
for pile batter and eccentricity of applied load. 

Computed distributions of load along the piles for the reanalyzed 
case are shown in Figs. 4.10 and 4.11 for applied group loads of 581.4 
kips (2.59 mN) and 1274.7 kips (5.67 mN), respectively. The largest 
deviations from measured behavior appear to be general underestimation 
of tip load by the model and failure of the model to replicate the high 
load (primarily high tip load) produced in the center pile at the failure 
load, which is represented in Fig. 4.11. 

It should be emphasized that the loads shown in Figs. 4.10 and 
4.11 are apparent loads that do not include residual loads that existed 
prior to loading. 

Subgroups . Analysis was made only for the reanalyzed case. 
Computed and measured load-settlement curves are shown for the two 
subgroup tests in Fig. 4.12. The PILGP1 soil inputs were exactly as 
for the 9-pile test, and the load-settlement curves displayed are for E 
(uniform) = 25 ksi (172 mN/m 2 ). Good agreement was achieved with 
these parameters for the 5-pile group. Settlements for the 4-pile group 
were slightly excessive, suggesting that E should have been slightly 
greater for this group. It is speculated that the requirement for a 
higher E for the 4-pile group may be the result of wider pile spacing 
(4.2 d compared with 3 d for the 9-pile group) and to the effects of 
prior loads to failure on the behavior of the 4-pile subgroup, which was 
tested after the 9-pile group and the 5-pile subgroup. The measured 
and best computed load-settlement curves for the 9-pile test are also 
shown in Fig. 4.12 for purposes of comparision. 

Computed and measured settlement ratios for the subgroups may be 
compared by referring to Fig. 4.9, and distribution of loads to the pile 
heads at load values representative of working loads may be seen in 
Table 4.4. Computed and measured load distributions along piles for an 
applied load of 278.9 k (1241 kN) on the 5-pile subgroup are plotted in 
Fig. 4.13. Figure 4.14 shows similar plots for an applied load if 287.6 
k (1280 kN) on the 4-pile subgroup. The computed relationships of 
load to depth deviate from the measured relationships in a manner 
similar to that observed for the 9-pile test. 

Observations 

The hybrid model, in the algorithmic form of Program PILPG1, 
appears to have yielded satisfactory results for this field test study 
when unit soil resistance curves developed from reference piles and 



186 



2 

LO 

•** 

II 

M 



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UJ 

co 

>- 



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-3 



2 
O 
PS 
tu 

CO 

pj 

hJ 
h- ( 

Cu 

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CO 



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pu 
o 

z 
o 

h- 1 
09 

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pj 

pa 
< 







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us 


r->. 


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00 


r^~ 


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3 












































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2 


to 


to 


LO 

to 


LO 

to 


to 


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<* 


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MEASURED 



FIGURE 4.14. MEASURED AND COMPUTED DISTRIBUTION OF LOADS ALONG PILES; 
4-PILE TEST; LOAD = 287.6 K (1 k = 4.45 kN; 1 ft = 0.305 m) 

192 



elastic soil properties representative of incompressible soil behavior 
(Poisson's ratio = 0.5) and of very low strain levels (constant Young's 
modulus approximately twice the value of the modulus measured by the 
self-boring pressuremeter at the level of the pile tips) are used. The 
authors realize that use of the term "satisfactory" implies an element of 
judgment on their part and that the reader may wish to interpret the 
term in light of specific criteria that he or she might have for assessing 
modeling accuracy. 

Use of criteria load transfer curves and ideal geometry also appear 
satisfactory, although less accurate. 






193 



Chapter 5. 
Recommendations for Future Study 

Based upon observations that the authors have made during this 
study, the following general recommendations for further research into 
the behavior of statically, vertical loaded pile groups are offered. 

1. While the study reported herein is believed to represent an 
important step in the understanding of pile group behavior, direct 
application of the experimental findings is limited, as discussed earlier. 
Further full-scale tests are necessary to develop a similar level of 
understanding in other soils. In particular, there exists a general 
understanding in other soils. In particular, there exists a general 
paucity of data for pile groups in sand, and the existing test data are 
contradictory with respect to factors needed by designers, including 
efficiency and settlement ratio. These apparent anomalies may be 
primarily the result of installation effects (predrilling, partial jetting, 
order of driving, installation of all piles simultaneously) and of the 
in-situ conditions of the sand (density, compressibility, degree of 
overconsolidation, piezometric conditions, stratigraphy). The 
"mechanical" interaction effect, that is, the settlement and load transfer 
induced in one pile in a group by loadings on other piles, is probably 
less important and can probably be handled adequately by existing 
analytical procedures (e.g., by one or more of the models described in 
the Interim Report ) once the effects of installation on load transfer and 
load- settlement for various individual piles in a group are known. 

Therefore, a full-scale field study should be undertaken to test 
individual instrumented piles within groups of various sizes and 
spacings, and for appropriate reference piles, in which installation 
techniques are varied. Sand stratigraphy variations should also 
considered. Such a study should yield practical, statistically 
significant information and would be cost-effective compared to 
conducting tests on complete groups of piles. Full-scale, or near 
full-scale, tests would be warranted because of difficulty in the 
physical modeling of certain important effects such as arching and grain 
crushing within the sand. In this regard the program of physical 
model testing at various scales now being undertaken by Mr. Carl Ealy 
of the FHWA should provide useful insight into the minimum size of 
piles required for such a field study. Measurement of residual stresses 
in the field should be emphasized because it is believed that the 
"critical depth" (depth at which ultimate unit side resistance ceases to 
increase linearly with depth) may be strongly dependent on the residual 
stress distribution and that such distribution may be considerably 
different in pile groups than in isolated piles. 

2. The effects of cyclic and long-term loading for groups in 
sands and in normally or slightly overconsolidated clays can best be 
studied by instrumentation and careful observation of in-service pile 
groups. In order for such studies to be effective, considerable 



194 



attention must be given to details of load and settlement measurements, 
particularly with respect to long term stability, and to coordination of 
activities with the bridge construction contractor. High quality soil 
compressibility data must also be obtained. 

3. The bearing capacity of the pile cap and the cap's effect on 
load-settlement response may be important for bridge foundations that 
can settle enough for consistent cap capacity to develop. It is the 
authors' opinion that at working load (low-settlement) levels, cap 
resistance is too unpredictable to be relied upon in design, but at 
settlements of greater than perhaps 1 in. (25.4 mm) cap resistance may 
become predictable. Several analytical studies (see Interim Report ) 
have been published regarding cap-soil interaction in pile groups. 
Experimental verification or modification of the analytical results could 
be developed best through physical (not full-scale) testing because 
numerous parameters, including method of preparation of the cap bearing 
surface, degree of cap overhang, moisture content changes in the 
surface soils, and effects of cyclic and vibratory loading should be 
systematically studied. Such studies can be justified in a practical 
sense only if FHWA's and other research into tolerable movements of 
structures indicates that settlements exceeding about 1 in. (25.4 mm) 
are acceptable in significant numbers of structures. 

4. Research into the development of purely theoretical 
approaches to the assessment of single pile and pile group capacity and 
load-settlement characteristics should be continued if installation of 
extremely long piles is contemplated for transportation-related 
structures . 

Behavior of such piles (e.g., piles longer than about 150 ft. (45.8 
m)), especially in groups, is largely beyond the limit of empirical 
knowledge. This may continue to be the case because of the expense 
involved in conducting full-scale load tests on instrumented piles of that 
length. A strong effort should be made to validate newly developed 
theories by comparing predictions to measurements acquired recently in 
several notable tests, including tests conducted for the FHWA at Ellis 
Island, Mo., the tests conducted at the Keehi interchange, Oahu, 
Hawaii, and the tests reported herein. 

5. With regard to Recommendation 4, future field test experi- 
ments should be designed to measure effective stresses against the 
faces of friction piles and at points within the soil mass. This 
recommendation is made because new completely theoretical models will 
likely involve the direct or implied usage of effective stresses. Before 
such measurements can be made reliably, further development of total 
stress and/or effective stress cells should be undertaken. 

6. No prediction of single pile or pile group behavior can be 
made reliably unless proper soil information is available. The results of 
this study and of other experiences by the authors suggest that soil 



195 



properties derived from in-situ testing procedures should lead to 
enhanced prediction of pile shaft and tip capacities and of deformations 
in the soil mass. Research into the development of in-situ test methods 
should therefore be continued with an emphasis on simple methods that 
will be implemented by potential users. 





w.cnnMENT PRINTING OFFICE: '98' ~34'-428/'202--' -3 



FEDERALLY COORDINATED PROGRAM (FCP) OF HIGHWAY 
RESEARCH AND DEVELOPMENT 



The Offices of Research and Development (R&D) of 
the Federal Highway Administration (FHWA) are 
responsible for a broad program of staff and contract 
research and development and a Federal-aid 
program, conducted by or through the State highway 
transportation agencies, that includes the Highway 
Planning and Research (HP&R) program and the 
National Cooperative Highway Research Program 
(NCHRP) managed by the Transportation Research 
Board. The FCP is a carefully selected group of proj- 
ects that uses research and development resources to 
obtain timely solutions to urgent national highway 
engineering problems.* 

The diagonal double stripe on the cover of this report 
represents a highway and is color-coded to identify 
the FCP category that the report falls under. A red 
stripe is used for category 1, dark blue for category 2, 
light blue for category 3, brown for category 4, gray 
for category 5, green for categories 6 and 7, and an 
orange stripe identifies category 0. 

FCP Category Descriptions 

1. Improved Highway Design and Operation 
for Safety 

Safety R&D addresses problems associated with 
the responsibilities of the FHWA under the 
Highway Safety Act and includes investigation of 
appropriate design standards, roadside hardware, 
signing, and physical and scientific data for the 
formulation of improved safety regulations. 

2. Reduction of Traffic Congestion, and 
Improved Operational Efficiency 

Traffic R&D is concerned with increasing the 
operational efficiency of existing highways by 
advancing technology, by improving designs for 
existing as well as new facilities, and by balancing 
the demand-capacity relationship through traffic 
management techniques such as bus and carpool 
preferential treatment, motorist information, and 
rerouting of traffic. 

3. Environmental Considerations in Highway 
Design, Location, Construction, and Opera- 
tion 

Environmental R&D is directed toward identify- 
ing and evaluating highway elements that affect 



* The complete seven-volume official statement of the FCP is available from 
the National Technical Information Service, Springfield, Va. 22161. Single 
copies of the introductory volume are available without charge from Program 
Analysis (HRD-3), Offices of Research and Development, Federal Highway 
Administration, Washington, D.C. 20590. 



the quality of the human environment. The goals 
are reduction of adverse highway and traffic 
impacts, and protection and enhancement of the 
environment. 

4. Improved Materials Utilization and 
Durability 

Materials R&D is concerned with expanding the 
knowledge and technology of materials properties, 
using available natural materials, improving struc- 
tural foundation materials, recycling highway 
materials, converting industrial wastes into useful 
highway products, developing extender or 
substitute materials for those in short supply, and 
developing more rapid and reliable testing 
procedures. The goals are lower highway con- 
struction costs and extended maintenance-free 
operation. 

5. Improved Design to Reduce Costs, Extend 
Life Expectancy, and Insure Structural 
Safety 

Structural R&D is concerned with furthering the 
latest technological advances in structural and 
hydraulic designs, fabrication processes, and 
construction techniques to provide safe, efficient 
highways at reasonable costs. 

6. Improved Technology for Highway 
Construction 

This category is concerned with the research, 
development, and implementation of highway 
construction technology to increase productivity, 
reduce energy consumption, conserve dwindling 
resources, and reduce costs while improving the 
quality and methods of construction. 

7. Improved Technology for Highway 
Maintenance 

This category addresses problems in preserving 
the Nation's highways and includes activities in 
physical maintenance, traffic services, manage- 
ment, and equipment. The goal is to maximize 
operational efficiency and safety to the traveling 
public while conserving resources. 

0. Other New Studies 

This category, not included in the seven-volume 
official statement of the FCP, is concerned with 
HP&R and NCHRP studies not specifically related 
to FCP projects. These studies involve R&D 
support of other FHWA program office research. 



DOT LIBRARY 



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